Applied Science, Faculty of
Civil Engineering, Department of
DSpace
UBCV
Dickof, Carla
Master of Applied Science - MASc
University of British Columbia
This paper examines CLT-steel hybrid systems at three, six, and nine storey heights to increase seismic force resistance compared to a plain wood system. CLT panels are used as infill in a steel moment frame combining the ductility of a steel moment frame system with a stiffness and light weight of CLT panels. This system allows for the combination of high strength and ductility of steel with high stiffness and light weight of timber. This thesis examines the seismic response of this type of hybrid seismic force resisting system (SFRS) in regions with moderate to high seismic hazard indices. A detailed non-linear model of a 2D infilled frame system and compared to the behavior of a similar plain steel frame at each height. Parametric analysis was performed determining the effect of the panels and the connection configuration, steel frame design, and panel configuration in a multi-bay system. Static pushover loading was applied alongside semi-static cyclic loading to allow a basis of comparison to future experimental tests. Dynamic analysis using ten ground motions linearly scaled to the uniform hazard spectra for Vancouver, Canada with a return period of 2% in 50 years as, 10% in 50 years, and 50% in 50 years to examine the effect of infill panels on the interstorey drift of the three, six, and nine storey. The ultimate and yield strength and drift capacity are determined and used to determine the overstrength and ductility factors as described in the National Building Code of Canada 2010.
CLT INFILL PANELS IN STEEL MOMENT RESISTING FRAMES AS A HYBRID SEISMIC FORCE RESISTING SYSTEM by Carla Dickof B.A.Sc., The University of Alberta, 2007 A THESIS SUBMITTED IN PARTIAL FULFILLMENT OF THE REQUIREMENTS FOR THE DEGREE OF MASTER OF APPLIED SCIENCE in THE FACULTY OF GRADUATE STUDIES (Civil Engineering) THE UNIVERSITY OF BRITISH COLUMBIA (Vancouver) April 2013 © Carla Dickof, 2013 ii Abstract This paper examines CLT-steel hybrid systems at three, six, and nine storey heights to increase seismic force resistance compared to a plain wood system. CLT panels are used as infill in a steel moment frame combining the ductility of a steel moment frame system with a stiffness and light weight of CLT panels. This system allows for the combination of high strength and ductility of steel with high stiffness and light weight of timber. This thesis examines the seismic response of this type of hybrid seismic force resisting system (SFRS) in regions with moderate to high seismic hazard indices. A detailed non-linear model of a 2D infilled frame system and compared to the behavior of a similar plain steel frame at each height. Parametric analysis was performed determining the effect of the panels and the connection configuration, steel frame design, and panel configuration in a multi-bay system. Static pushover loading was applied alongside semi-static cyclic loading to allow a basis of comparison to future experimental tests. Dynamic analysis using ten ground motions linearly scaled to the uniform hazard spectra for Vancouver, Canada with a return period of 2% in 50 years as, 10% in 50 years, and 50% in 50 years to examine the effect of infill panels on the interstorey drift of the three, six, and nine storey. The ultimate and yield strength and drift capacity are determined and used to determine the overstrength and ductility factors as described in the National Building Code of Canada 2010. iii Preface I created and calibrated all the analytical models and wrote all of the contained manuscript. The material models developed in Chapter 5 were based on the experimental tests performed at FP Innovations by Johannes Schneider for his 2009 thesis, (Connections in Cross- Laminated-Timber Shear Walls Considering the Behaviour under Monotonic and Cyclic Lateral Loading 2009). I was responsible for the calibration of the material model to match the experimental results as well as all the development of the system model and the resulting analytical results discussed in Chapter 6. This thesis is made on the topic the effect on strength and seismic behavior of a steel moment frame with CLT infill panels. Versions of chapters 2 through 6 have been published in the following conference papers: • Stiemer, Dickof, Tesfamariam, “Wood-Steel Hybrid Systems: Overstrength and Ductility in Design”, 10th International Conference on Advances in Steel Concrete Composite and Hybrid Structures, Singapore, 2012 • Stiemer, Dickof, Tesfamariam, “Steel-Timber Hybrid Structures – Design Performance and Dynamic Behaviour”, 22nd Australasian Conference on the Mechanics of Structures, Sydney Australia, 2012 • Dickof, Stiemer, Tesfamariam, “Wood-Steel Hybrid Seismic Force Resisting Systems: Seismic Ductility”, World Conference on Timber Engineering, Auckland NZ, 2012 The research conducted in the above publications and this thesis was completed by me, with guidance from Dr. Siegfried Stiemer and Dr. Solomon Tesfamariam. iv Table of Contents Abstract .................................................................................................................................... ii Preface ..................................................................................................................................... iii Table of Contents ................................................................................................................... iv List of Tables ......................................................................................................................... vii List of Figures ....................................................................................................................... viii Acknowledgements ................................................................................................................. x Chapter 1: Introduction ........................................................................................................ 1 Chapter 2: Materials.............................................................................................................. 4 2.1 Connections............................................................................................................... 5 2.2 Crossed Laminated Timber (CLT)............................................................................ 6 2.2.1 Experimental Tests................................................................................................ 7 FP Innovations (Popovski and Karacabeyli 2011)........................................................ 7 Ceccotti et al. (2010) ................................................................................................... 10 Fragiacomo et al. (2011) ............................................................................................. 11 2.2.2 Analytical Models of CLT Systems .................................................................... 12 Rinaldin, Amadio and Fragiacomo (2011) ................................................................. 12 Ceccotti (2008)............................................................................................................ 13 Chapter 3: Hybridization .................................................................................................... 14 3.1 Hybrid System Case Studies ................................................................................... 15 Kanazawa M Building (Koshihara, H. and Yusa 2005) ............................................. 15 Scotia Place (Moore 2000) .......................................................................................... 17 Quebec Six Storeys Hybrid Office Building (Mohammad, et al. 2011) ..................... 18 v 3.2 Infill Shear Walls .................................................................................................... 19 Chapter 4: Seismic Force Resisting System Design .......................................................... 21 4.1 Ductility and Force Based Design .......................................................................... 21 4.2 Performance Design ................................................................................................ 23 Chapter 5: Methodology...................................................................................................... 25 5.1 Sample Structure ..................................................................................................... 25 5.2 Design and Modeling .............................................................................................. 28 5.2.1 OpenSees............................................................................................................. 29 5.2.2 Frame Design and Modeling ............................................................................... 30 5.2.3 Wall Design and Modeling ................................................................................. 33 5.2.4 Connection Design and Modeling ...................................................................... 34 5.3 Loading ................................................................................................................... 39 5.3.1 Semi-Static Monotonic Analysis ........................................................................ 40 5.3.2 Semi-static Cyclic Analysis ................................................................................ 40 5.3.3 Ground Motion Analysis..................................................................................... 41 Chapter 6: Results................................................................................................................ 44 6.1 Single Bay, Single Storey, Static and Semi-Static-Cyclic Results ......................... 44 6.2 Multi-Storey Static Results ..................................................................................... 50 6.2.1 Ductility Factors.................................................................................................. 58 6.2.2 Overstrength Factors ........................................................................................... 61 6.3 Seismic Response.................................................................................................... 64 Chapter 7: Conclusion ......................................................................................................... 68 7.1 Summary of Findings .............................................................................................. 68 vi 7.2 Future Research ...................................................................................................... 70 References .............................................................................................................................. 71 Appendices ............................................................................................................................. 75 Appendix A Frame Design ................................................................................................. 75 Appendix B Scaled Ground Motions .................................................................................. 80 B.1 Ground Motions .................................................................................................. 80 B.2 Scaling Factors .................................................................................................... 80 vii List of Tables Table 2.1: Materials Properties: Steel, Wood, and Concrete (Khorasani 2010) ....................... 4 Table 4.1: Steel Ductility and Overstrength Information ....................................................... 23 Table 5.1: Single Bay Parameters of Interest .......................................................................... 27 Table 5.2: Multi-Storey Parameters of Interest ....................................................................... 28 Table 5.3: Rotation Requirements for Plastic Hinges (American Society of Civil Engineers (ASCE) 2006) ......................................................................................................................... 33 Table 5.4: CLT Material Properties (Structurlam n.d.) .......................................................... 33 Table 5.5: Amplitude for Semi-static Cyclic CUREE Protocol ............................................. 36 Table 5.6: Amplitude for Semi-static Cyclic ISO Protocol .................................................... 41 viii List of Figures Figure 2.1: Crossed Laminated Timber (CLT) ......................................................................... 7 Figure 2.2: Hysteretic Response of CLT Walls to Lateral Loading (Schneider 2009) ............. 8 Figure 2.3: Semi-static CLT Wall Tests - Effect of Connection Between Panels (Popovski and Karacabeyli 2011) .............................................................................................................. 9 Figure 2.4: 7 Storey CLT Shake Table Test (Fragiacomo, Dijic and Sustersic 2011) ........... 11 Figure 3.1: Brace Connection Schematic................................................................................ 16 Figure 3.2: Floor Assembly Schematic ................................................................................... 17 Figure 3.3: Six Storey Wood and Concrete Hybrid Building Structure (Mohammad, et al. 2011) ....................................................................................................................................... 18 Figure 3.4: Column Connections ............................................................................................ 19 Figure 5.1: Base Building Floor Plan (left); Base 2D Frame Elevation (right) ..................... 26 Figure 5.2: Single Bay, Single Storey, CLT Infilled Frame with Bracket Locations ............. 27 Figure 5.3: Wall Locations for Frame with Infill Wall Locations .......................................... 28 Figure 5.4: Modified Ibarra Krawinkler Deterioration Model for BILIN Material in OPENSEES (Lignos and Krawinkler 2011) ........................................................................... 32 Figure 5.5: Failed Bracket (left); Bracket Stress-Strain Curve (right) (Schneider 2009) ....... 34 Figure 5.6: Two Node Link Formulation (Schellenberg n.d.) ................................................ 35 Figure 5.7: Comparison of Test and Analytical Bracket Behavior ......................................... 36 Figure 5.8: Test and Analytical Bracket Behavior Comparison of Important Values ............ 37 Figure 5.9: Elastic Perfectly Plastic Gap Material Behavior (Schellenberg n.d.) ................... 38 Figure 5.10: Combined Bracket Material ............................................................................... 39 Figure 5.11: Static Pushover Initial Periods ........................................................................... 42 Figure 5.12: Scaled Ground Motion Spectrums Compared to Vancouver’s Hazard .............. 43 ix Figure 6.1: Effect of Infill CLT Panel .................................................................................... 45 Figure 6.2: Effect of Wood Crushing Strength ....................................................................... 46 Figure 6.3: Force in Bracket vs. Overall System Drift for all Brackets Along Both Columns ................................................................................................................................................. 47 Figure 6.4: Effect of Confinement Gap on a Frame with Infill Panel .................................... 48 Figure 6.5: Cyclic and Monotonic Behavior of Single Bay Infilled Frame. Infilled Frame with 0.0mm Gap (left); Infilled Frame without Confinement (center); Bare Moment Frame (right) ...................................................................................................................................... 49 Figure 6.6: Parametric Monotonic Pushover Curves .............................................................. 51 Figure 6.7: Ultimate Strength Compared to The Number of CLT Infilled Bays .................... 52 Figure 6.8: Ultimate Drift Capacity Compared to the Number of CLT Infilled Bays ............ 53 Figure 6.9: Yield Strength Compared to the Number of CLT Infilled Bays .......................... 55 Figure 6.10: Yield Drift Capacity Compared to the Number of CLT Infilled Bays ............... 57 Figure 6.11: Ductility Factors for Systems with Type D Frames ........................................... 58 Figure 6.12: Ductility Factors for System for Type LD Frames ............................................. 60 Figure 6.13: Overstrength Factors for System for Type D Frames ........................................ 62 Figure 6.14: Overstrength Factors for System for Type LD Frames ...................................... 62 Figure 6.15: Nine Storey Infilled Ductile Frame Building Drift at Ultimate Drift ................. 65 Figure 6.16: Nine Storey System with Ductile Frame Maximum Inter-storey Drift ............. 65 Figure 6.17: Nine Storey Inter-Drift for all Frame-Infill Configurations ............................... 66 Figure 6.18: Six Storey Inter-Drift for all Frame-Infill Configurations ................................. 67 x Acknowledgements I am grateful for all the support of the faculty, staff, and my fellow students at UBC. I would particularly like to thank my supervisors, Dr. Siegfried Stiemer and Dr. Solomon Tesfamariam, for their guidance. The feedback and support from the industrial collaborators, Dr. Marjan Popovski and Mr. Erol Karacabeyli from FP Innovations, is gratefully acknowledged. Special thanks to Mr. Johannes Schneider for his input and collaboration. This research was supported through funding to the NSERC Strategic Network on Innovative Wood Products and Building Systems. Last but not least, I would like to thank my parents for their unending support. 1 Chapter 1: Introduction Hybrid systems are found throughout the world, in many types of structures, and with many different materials. While any system that combines two or more materials to resist loading can be defined as hybrid; using steel and concrete together is the most common in modern construction. This includes concrete on metal deck supported on steel beams as a floor system as well as the typical concrete reinforced with steel. Steel and timber hybrid systems are less common, but they do exist. For example, Quebec and Northern Ontario have many steel and wood hybrid bridges (Krisciunas 1996). Some steel and timber systems use timber as the secondary structure, such as floor joists, with the primary system constructed from steel, such as the columns, primary beams, and braces. Hybridization can be divided into three categories: component level, system level, and building level. In general, the goal of each is to take advantage of the strength of each material while reducing the impact of their weakness. The main focus of this thesis is on system level hybridized system using wood and steel; specifically a vertical seismic force resisting system (SFRS) combining steel moment frames with Cross-Laminated-Timber (CLT) shearwall panels. Chapter 2 reviews the material properties of wood and steel as well as highlighting potential incompatibilities. Steel is much stronger and provides significant post-yield deflection capability, and so moment frames of steel are extremely ductile, but they generally experience large deflections during seismic events. Wood is comparatively much weaker larger members are usually required, resulting in stiffer systems. Furthermore, wood does not have significant deformation capacity post-yield, especially when loaded perpendicular to the grain, resulting in a less ductile system overall. The material incompatibilities are overviewed as they pose important problem for the connections of this kind of system. Steel connectors are commonly used in timber structures and are effectively a small scale hybridized system. A discussion of typical wood connections is provided in this section as well. Finally, CLT is introduced along with the common connectors used in a plane CLT building. Significant experimental testing has been completed on CLT wall panels and analytical models have been created for CLT panel systems; the results are summarized herein. 2 Chapter 3 describes component, system, and building level hybridization while provided specific examples of each. Three case studies of existing steel and timber hybrid buildings are presented, including one six storey building for Canada. The issues that arose in the case studies presented are discussed and the advantages are presented. Finally, the type of hybridization that will be the focus of this thesis is discussed. Steel moment frames with infill CLT panels is not a system that has notable documentation or case studies, but in many ways it is similar to masonry infilled moment frames. That type of system is seen in both steel and several studies are presented demonstrating significantly decrease the drift of the structures as well as increasing the ultimate and yield strength of the system. These effects should translate directly to infill CLT panels Chapter 4 describes the equivalent static force design method in the National Building Code of Canada (NRC 2010) as well as the method for developing the ductility and overstrength factors. The force exerted on the building due to an elastic response is calculated, and then reduced by the ductility factor (µ, Rd) and the overstrength factor (Ω, Ro) to determine the forces that must be resisted after yielding has occurred. Other design codes provide similar reduction factors for force based design. Chapter 5 discusses the parametric analyses performed in OpenSees to determine the characteristics of a hybrid steel moment frame with infill CLT shearwall panels. The steel frame is created with non-linear hinge lengths at the member ends to provide non-linear behavior and the CLT infill panels are simplified as linear-elastic isotropic quadrilateral shell elements that experience stress only in plane. The connectors between the timber shell elements and the steel frame elements are modeled as non-linear links; the material model associated with the links combines the elastic-plastic behavior of the panel crushing with the pinching hysteretic behavior of a nailed bracket connection. Chapter 6 reviews the static pushover analysis and dynamic ground motion analysis are performed in the parametric study. The pushover analysis was done for the full spectrum of parameters, which include steel moment frames ductility, panel thicknesses, panel strength, and connection configurations. The pushover analyses are used a basis to provide initial estimates of ductility and overstrength for the hybrid system. Dynamic analysis was 3 performed using 10 linearly scaled ground motions scaled to Vancouver’s hazard spectrum at 2% in 50 years, 10% in 50 years, and 50% in 50 year return periods. The results of these ground motion tests are used to analyze the interstorey drifts and associated base shears to help refine the ductility and overstrength factors. In Chapter 7, conclusions and future direction are provided. 4 Chapter 2: Materials To optimize combination of different materials in a hybrid structures, with consideration of their respective advantages and compensate for their respective disadvantages, a full understanding of the material properties is required. The characteristics of wood and steel are summarized in Table 2.1. Table 2.1: Materials Properties: Steel, Wood, and Concrete (Khorasani 2010) Material properties Steel Timber Concrete Density (kg/m3) 7800 400-600 2300 Modulus of Elasticity (MPa) 200 000 8000-11000 20 000 Strength (MPa) Compression 400 - 1000 Parallel: 30 Perp: 8 20-40 Tension 400 - 1000 Parallel: 6 Perp: 1 2.0-5.0 Yield 350 N/A N/A Steel has much more significant post-yield deformation capacity, or ductility. Additionally, wood has much more significant post yield behavior for compression forces compared to tension forces. Steel is an isotropic material, meaning that its characteristics are constant throughout the material. It also has the largest unit mass of any of the typical construction materials. Because of its extremely high strength, the challenge with steel is to use the least amount of material possible, resulting in buckling governing design of columns and compression zones in beams. As a result, steel is the ideally used in tension. Unlike concrete and steel, wood is an anisotropic material. This means that the material has different strengths depending on the orientation and location where the load is applied. Wood has different strengths in its three axes: longitudinal, radial, and tangential. Wood is strongest parallel to the grain, or longitudinally, and weakest perpendicular to the grain in the radial direction specifically. Generally the species and specific characteristics of the tree the timber was harvested from changes the strength of the material, therefore wood has significantly more variable than other traditional building materials, steel and concrete. Timber characteristics are extremely dependent on the species of tree and specific qualities of the wood harvested; growing conditions can have a large impact as well as local imperfections in the wood, such as knots 5 (Keenan 1986). Another important weakness to note in any use of timber is its susceptibility to rolling shear, shear leading to strain perpendicular to the grain. If wood is thought of a series a parallel fibers held together with fairly week bonds between them, then rolling shear results in the fracture of this type of bond and the fibers “rolling” next to one another. It tends to cause detachment of one layer of wood grain from another. Wood is much weaker in this type of shear than in “radial” or “longitudinal” shear. Also, it is important to note the comparative hydroscopic and thermal properties of wood compared to steel. Steel is very thermally sensitive, expanding and contracting as temperatures rise and fall respectively, wood is thermally very stable. Comparatively, wood is a hydroscopic material, expanding and contracting as its moisture content rises and falls, respectively, moisture has no effect on the dimensions of steel. The hydroscopic properties of wood are also affected by its anisotropic properties. The tangential direction shows the larger shrinkage, followed by the radial direction, with the longitudinal direction showing very limited shrinkage effects. Tangential shrinkage is generally approximately twice that of radial shrinkage, although this varies depending on the species. Tangential shrinkage from green to oven dry ranges from 5% to 12.5%; comparatively, radial shrinkage ranges from 2.2% to 7.7% depending on species. The different amount of shrinkage in different directions can cause the warping noticeable in some old pieces of wood; this is dependent on how the tree was milled. Longitudinal shrinkage is almost negligible with most species of wood showing 0.1 to 0.2% shrinkage longitudinally. 2.1 Connections Hybrid system should be designed by taking advantage of each material. For this reason, timber seems ideal to replace typical floor systems and/or wall systems (e.g. shear walls). This can significantly reduce the weight compared with concrete or masonry infill. There are significant differences between the ways the materials respond to the load, and this should be taken into consideration. Timber shear walls respond very differently than timber braces or timber shear walls: Timber frames should be designed with different goals that steel frames. Steel frames are designed 6 generally to be more ductile. Timber frames cannot achieve the same ductility. Ductility is typically introduced into a timber system through deformation of steel connectors. Connections are important. They must be resolved between the wood members and between the timber-steel interfaces. One such example is the use of steel dowels in the form of bolts, nails, rivets, etc. As the dowels deform plastically, they also cause of local crushing of the wood at the connection location. An ideal connection involves the largest amount of yielding possible, with the minimum amount of wood crushing. Steel has the largest deformation capacity and can dissipate large amounts of energy, a desirable property during a seismic event. Wood crushing also contributes a significant amount of post-yield displacement capacity. Further, as crushing occurs and the wood density increases, the strength of the wood increases both perpendicular and parallel to the grain. Wood splitting should be avoided; it results in brittle failure. Steel connection plates are commonly used to connect timber members together, or to connect timber to steel elements. Steel plate connection brackets are common in many steel- timber hybrid buildings. Further, connections must also account for different material type incompatibilities. For example, wood shrinks and swell with changing moisture types; steel expands and contracts with temperature changes. Care must be taken to allow for these material size changes in the connection so as not to place unintended internal stresses in the connections. Sometimes this is as simple as providing slotted bolt holes in the steel. The expansion/contraction and shrinkage/swelling of the materials can be most sensitive during construction before the building envelope is in place, which helps control the temperature and moisture of the air surrounding the structure. 2.2 Crossed Laminated Timber (CLT) Both shear walls and braced frames are commonly used wood lateral systems, similar to other common building materials such as steel and concrete. The most common type of wood shear wall is the sheathed stud wall: studs with nailed plywood or Oriented Strand Board (OSB). This thesis will focus on a new type of wood shearwall panel – Crossed 7 Laminated Timber panels. Crossed laminated timber is layered and glued similar to typical glued laminated beams. The only different is the orientation of the planks used; crossed laminated timber (CLT) attempts to approximate an orthotropic plate. A cross section of CLT is shown in Figure 2.1. Figure 2.1: Crossed Laminated Timber (CLT) 2.2.1 Experimental Tests Determining the design properties of CLT is not necessarily easy. It depends on the species and quality of wood used as well as the number, orientation, and thickness of the layers. Several experimental tests have been run on CLT wall systems ranging from component tests, to wall tests, to full building tests. Summary of each of these reported studies are provided below. FP Innovations (Popovski and Karacabeyli 2011) Several semi-static tests have been performed. The tests included variations on a wall panel with connectors at the base of the wall. Cyclic loading was then applied to the assembly. The base assembly included a 2.3m × 2.3m panel, in accordance with the CUREE testing protocol, with four connectors along the base of the panel. Two type of CLT walls responses are observed, overturning or rocking shearing or a combination of the two. 8 Typically rocking and deflection of the connections has the largest contribution to energy dissipation and is where failure occurs in the system. The resultant behavior of the walls shown in Figure 2.2 shows hysteretic pinching behavior. The base connections have a significant impact on the system; the deformation of the panel itself is only a portion of the lateral behavior of CLT walls, and is frequently approximated to zero when modeling wall systems. Figure 2.2: Hysteretic Response of CLT Walls to Lateral Loading (Schneider 2009) The hysteretic response clearly shows an ultimate strength of 90kN or 39kN/m and a yield strength of approximately 75kN or 30kN/m. The initial stiffness of the system is 3.5kN/m which includes the slip due to the connections. It is important to note that additional gravity load increases the strength and stiffness of CLT walls. Further, connections at joints between wall panels can have a large impact on the behavior of the wall. Deflection in these fasteners can increase the ductility in the system; comparatively, the peak strength of the system decreased with an increase in step joints between panels. As shown in Figure 2.3. 9 Si n gl e Pa n el CL T sh ea r w al l Th re e pa n el CL T sh ea r w al l; sc re w co n n ec tio n be tw ee n pa n el s Figure 2.3: Semi-static CLT Wall Tests - Effect of Connection Between Panels (Popovski and Karacabeyli 2011) 10 Ceccotti et al. (2010) Based on testing done in Europe and recommendations for the Eurocode ductility factor, q, recommendations have been made for conservative ductility (Rd) and overstrength (Ro) factors for the NBCC. According to evaluation at FPInnovations in Vancouver, Rd = 2.0 and Ro = 1.5 are conservative estimates for pure CLT structures with nailed and screw connections. Further, the behavior is superior to that of braced timber frames given then same factors as CLT constructions “is not susceptible to the soft storey mechanism as the panels (that are also vertical load carrying elements) are virtually left intact in place even after a “near collapse” state is reached” (Popovski and Karacabeyli 2011). Large scale dynamic tests have also been performed on full assemblies. A three storey CLT building test was performed on a unidirectional shake table by NIED and CNR-IVALSA in Japan. Tests were performed using Kobe, El Centro, and Nocera Umbra ground motions adjusted to peak ground accelerations for 0.15g and 0.5g. Finally, the ground motion acceleration was increased until the “near-collapse” state. The test specimen was approximately 7m × 7m in plan and 10m tall. The walls were composed on 85mm thick wall panels and 142mm thick floor panels (Ceccotti, Sandhaas and Yasumuro 2010). No damage was observed in any component at a peak ground acceleration of 0.5g. When the ground acceleration was increased to a maximum of 0.8g, slight deformation was noticed in the screws at the vertical joints between the panels, but no other damage was observed. Finally, when the peak ground acceleration was increased to 1.2g, hold down failure was observed through pull out and bending of the nails; deformation in the screws between the panels was also observed. 11 Fragiacomo et al. (2011) A seven storey full scale CLT building was tested at the E-defense shake table in Japan. This shake table allows for ground motions to be applied in three dimensions simultaneously. Figure 2.4: 7 Storey CLT Shake Table Test (Fragiacomo, Dijic and Sustersic 2011) The seven storey building was subjected to 100% of the Kobe earthquake ground motion with a peak ground acceleration of 0.82g in one direction and 0.6g in the perpendicular direction. The building responded with limited structural damage. Some damage to the connectors in the hold downs were noticed, although no failure occurred. Additionally, with appropriate ductility connections between wall panels, an Rd of 3.0 is achieved (Fragiacomo, Dijic and Sustersic 2011). 12 2.2.2 Analytical Models of CLT Systems When CLT is modeled in a wall or diaphragm system, the CLT is often modeled as a rigid panel, or in some cases a simplified rigid frame, and the connections are modeled as non- linear line elements. Several studies show this same trend. Rinaldin, Amadio and Fragiacomo (2011) Rinaldin et al. (2011) created a model of a single CLT panel with connecting brackets in ABAQUS. Similar to other analysis models, two non-linear hysteretic behaviour models were created for the brackets. One in shear only, to represent the angle bracket connections, and one for the tension and compression in the hold-down connections. The bracket spring model was defined with fourteen parameters including three parameters to determine the stiffness at various points, yield and peak strength, ultimate displacement, and five degradation characteristics. Comparatively, the hold-down spring model is non- symmetric, the behavior in compression is linear elastic with a very high stiffness to approximate contact between the CLT panel and the foundations. The input parameters for the tension behavior are similar to that of the bracket spring model. The input values were calibrated to experimental data from wall tests at Ivalsa CNR. Calibration of the input parameters was aided using So.ph.i (Software for Phenomenological Implementation). The timber panel was modeled as a shell element. The cross section of the shell was defined as five layers of linear elastic orthotropic wood material; similar to other analytical models, the assumption was that all plastic deformations would occur in the connectors. Three bracket spring models were placed along the base of the shell element, and hold-down springs were placed at the edges. Contact springs were also placed at the base of each shell along the bottom of the wall. Contact springs are specialized no tension, high compression stiffness springs. The results from this model were compared with the results from wall tests and were found to match closely in hysteretic behavior as well as total energy dissipated. 13 Ceccotti (2008) Ceccotti (2008) performed an analysis to predict the results of the 3D three storey building shake table test. The analytical model was created in Drain3D. The program was modified to allow for the type of non-linear behavior present in timber connections. The model consists of three major components, rigid panels modeled as stiff braced frames, and two types of non-linear springs. One type to represent the angle brackets, with symmetric non- linear pinching behavior to match experimental data. The other type of non-linear spring represents the hold-downs, with non-symmetric behavior; non-linear pinching behavior is modeled in tension and very stuff linear elastic behavior is modeled in tension. The Nocera Umbra ground motion from the experimental shake table test scaled to a peak ground acceleration of 1.2g was applied to the analytical model; the results were found match well with the test results from the shake table test. For the Kobe earthquake ground motion, scaled to 0.82g peak ground acceleration, the uplift shown in the modeled varied a maximum of 6% from the experimental data. As the peak ground acceleration increased, the difference between the model and the experimental data increased. For the failure case, a 1.2g peak ground acceleration on the Nocera Umbra ground motion, a maximum of 40% difference between the modeled response and the experimental results was shown (A. Ceccotti 2008). 14 Chapter 3: Hybridization A hybrid building uses two or more materials in combination to get the most out of each material. This thesis focuses on steel and timber. Technically all timber buildings are hybrid systems as the connections are almost always made of steel (nails for example). Hybridization can be broken down in to three types: Component Hybridization, System Hybridization, and Building Hybridization. Component level hybridization involved more than one material type within a member. One common example of component level hybridization for timber-steel hybrid is a flitch beam, composed of one or more steel plate(s) sandwiched between pieces of timber. There are several advantages to this type of system. The steel beam has significantly higher strength than the timber members, but a steel plate will generally have significant issues with lateral torsional buckling. The timber members provide lateral restraint for the steel preventing lateral torsional buckling or local buckling. The steel and wood are connected together using bolts spread over the length of the beam to transfer shear. This connection is the important part of the design to ensure that appropriate distribution of forces occurs without splitting in the wood. System level hybridization entails use of multiple material types within a structural system. The connections between these members are frequently the most complex issue in system level hybridization. A common example of a hybrid system is steel and wood trusses. Timber trusses can have distinct advantages for use in buildings where the stored materials may give rise to environments that could be corrosive to steel. In such situation, the amount of steelwork can be kept to a minimum so that expensive anti-corrosion costs are minimised. Typically these are constructed with a timber top chord for the truss and steel bottom chord. This works to each material’s advantages as timber has high compression strength and steel is best used in tension. Building level hybridization is the combination of building systems of different materials. One example of this is a vertically mixed system. Vertically mixed systems have been completed around the world, with one timber vertically mixed system in Australia recently becoming the tallest timber residential building in the world (Harris 2012). The lower 15 floor(s), which support more load and often have a higher storey height, are constructed entirely from concrete or steel framing; the storeys above are then timber framed. The result is a significantly reduced total weight of the structure, decreasing the size of the foundations as well as the requirements for the lower storey(s) design. The stronger material type at the lower floor(s) prevents the requirement for larger timber sizes at these levels. 3.1 Hybrid System Case Studies Although the combination of steel and wood in buildings is not as common as hybrid steel and concrete, there are still some reported studies. Two such case studies are reviewed here, Kanazawa M Building and Scotia Place. Kanazawa M Building (Koshihara, H. and Yusa 2005) An existing building in Japan, the Kanazawa M Building, is one case study of a hybrid building (Figure 3.1). It combines several kinds of hybridization both at the component and system level. The first storey is a reinforced concrete shear wall system; the second to fifth storeys are a steel and timber hybrid frame braced system. The beams, columns, and braces are all glued laminated timber with built in steel. The columns and braces are timber with solid square steel rods running through the center. The beams are wood with a central steel plate, or a flitch beam as described previously. The floors are concrete slabs; they are connected into the wood beams with steel plates in the concrete and lag screws into the beams. Standard plywood shear walls are also present in the building. The wood in the beams provides to functions: stiffening and strengthening the beam in bending under gravity loads, as well as preventing buckling due to compression under lateral loads, and lateral torsional buckling under gravity loads. The timber in the columns acts as fire-proofing as well as buckling restraint; a 3mm clearance is provided to prevent the timber from taking any of the gravity load accommodating the hydroscopic and temperature incompatibilities between the timber and steel. Testing showed the steel yielding in force compression instead of buckling when the timber was present. Similarly, the wood in the brace members serves to restrain the braces against buckling. The steel used is a low strength with a yield of 284 MPa; testing showed 16 the steel beginning to yield, along with some strength hardening. As the steel deformed, the timber came in contact with jig and began to take axial load; as load increased, eventually the timber cracked and buckled. The intent of the timber braced frame system is shown in Figure 3.1 Figure 3.1: Brace Connection Schematic The plywood shear walls in the other direction are constructed from 24mm plywood and 8mm diameter screws. The plywood was screwed to the timber in the hybrid columns and beams. At the base of the second floor, the wall connects into the concrete. Here the shear is transmitted through the screws into a wood sill which is then bolted to the concrete. 17 Scotia Place (Moore 2000) This is a building In New Zealand constructed from steel concentrically braced frames and wood floors on steel studs. The wood floor acts as the slab for gravity and diaphragm loads. The vertical structural system in this building is purely a steel frame system. A schematic of the wood deck at the primary steel framing members are shown in Figure 3.2. Figure 3.2: Floor Assembly Schematic The use of wood floors, managed to significantly reduce the weight of the building. This had significant effect on the structure for both gravity and lateral load resistance. The steel columns, concrete basement columns, and piles on which the building is supported were all made smaller as a result of the use of wood floors instead of concrete. Further, the reduction of weight in the structure resulted in the building frames being governed by wind loading. Only the design of the wood diaphragm was governed by the seismic loading. 18 Quebec Six Storeys Hybrid Office Building (Mohammad, et al. 2011) Building hybridization is the combination a different material system types within a building. One example that has been constructed in Canada is a six storey hybrid building built in Quebec. The system is heavy timber framing with concrete shear walls as shown in Figure 3.3. Figure 3.3: Six Storey Wood and Concrete Hybrid Building Structure (Mohammad, et al. 2011) This type of system has several advantages compared to an all concrete solution. The wood framed solution is significantly lighter than a concrete column and slab. This will reduce the total weight of the building, which will reduce the foundations and seismic design load (i.e. requirements on the shear walls); often resulting in the building being governed by wind design. Significant wind studies were done to determine the frequency and damping ratios of the first three modes of vibrations. Another important factor in the design was to prevent any issues due to the difference in shrinkage between the heavy timber frame and the concrete cores. If shrinkage affects the overall height of the timber frame, there would be a difference between the openings in the elevator and stair cores, and the floor elevations. To prevent this, the timber frame was designed with without any intermediate members between the columns. This effectively prevents any tangential or radial shrinkage affecting the height of floor elevations. In other 19 words connections between columns were designed to bear only on the end grain of other columns as shown in Figure 3.4. Figure 3.4: Column Connections 3.2 Infill Shear Walls This thesis will focus on one type of system hybridization: wood infill shear walls in a steel moment frame. This type of hybridization has been done before using other materials, most commonly masonry infill walls in either steel (reference) or concrete (reference) moment frames. Infill masonry walls are installed in steel and concrete moment frames. They effectively stiffen and strengthen the moment frame considerably. The design of these types of systems are complex; the walls “behave as a constituent part of the structural system and determines the overall behavior of the structure, especially when it is subjected to seismic loads” (Kodur et al. 1995). Many studies have been done to show the benefits of masonry infill walls on moment frame buildings including increased strength, stiffness, energy dissipation, and resistance to incremental collapse. There are problems that need to be addressed for this type of infill shear wall; the ductility of the system is significantly reduced by their inclusion in the structure. Unreinforced masonry walls are typically assigned a ductility ratio, Rd = 1.0. Masonry infill walls are typically represented and designed as a diagonal compression strut. Another type of infill wall is typical in construction; stud wall partitions are common in construction of all varieties. Typically in steel and concrete buildings, these infill partitions are not accounted for in the structural design except to account for their contributing mass. 20 Steel frame structures are very flexible; the high comparative stiffness of the infill non- structural partitions has been shown to have a significant impact on the seismic response of the structures due to the high comparative stiffness of the infill partitions despite being non- structural components. Yousuf and Bagchi’s (2009) study on the impact of non-structural infill partition walls on ductile (type D) steel moment frames found that the impact of infill partition walls results in decreased deflection and ductility in the structure. Further, hinging occurred in the columns at locations other than their base. The study showed that it is important to account effectively isolate non-structural partitions from moment frame system. Alternatively, partitions could be designed as structural items accounting for the strength of these walls and provide some guidelines on their constructions. 21 Chapter 4: Seismic Force Resisting System Design Seismic force resisting system (SFRS) design is generally done in one of two ways: force based design and displacement based design. Force based design quantifies maximum force (base shear) at the base of the structure, and then distributes this along the height of the building to determine the force at each storey, the storey shear; additionally the overturning moments are determined. The building is then designed to resist these maximum shears and moments. These forces are then reduced by the ductility and over strength factors. Force based design is the most common approach used in design codes today. Displacement based design, on the other hand, determines the maximum displacement that the structure will observe and then design the structure to resist this either elastically or plastically (Chopra 2011). Generally plastic design is used as this will dissipate energy during a seismic event. This will result in much larger amounts of deformation; acceptability criteria must then be assigned to estimate the amount of allowable damage that can occur with the building maintaining its required performance. This type of design is much newer; it is called performance based design and is only present in some design codes. In some cases, performance based design is still combined with force based design to allow designers to use familiar methods. 4.1 Ductility and Force Based Design Design Codes generally use a force based approach in the design of SFRSs. This method involves determining the force the system would experience if it were to remain elastic and then reduce it by some factor accounting for ductility, redundancy, and overstrength (Elnashai and Mwafy 2002). The focus of this research is the National Building Code of Canada 2010 which divides these into two factors: ductility factor (µ, Rd) and overstrength factor (Ω, Ro). The equivalent static force procedure (NRC 2010) calculates the base shear () based on the elastic spectral response adjusted by the ductility and overstrength factors as well as an importance factor () and a factor to accound for higher mode effects () shown in equation (1). 22  =    (1) where the spectral acceleration is defined based on the sites soil factor (, ) and the areas response spectrum   of the area at the buildings specific fundamental period (  .  =      < 0.5smin [   ,    ] 0.5s ≤  < 1.0s    ≥ 1.0s # (2) Mitchell et al. (2003) highlighted that the NBCC defines the system overstrength from a combination of factors including overstrength of the yield strength of the system ($%), the probable overstrength of the material (&), the overstrength due to strain hardening of the material ('), and the redundancy of the collapse mechanism (()'), and the actual yield to collapse overstrength (*+$,) which will be the focus of this paper.  = $%&+$,'()' (3) FEMA P695 (2009) outlines a detailed procedure for the development of a systems overstrength and ductility for the US design codes. The first factor Overstrength (Ω) is defined as the ratio of the maximum base shear resistance (Vmax) and the design base shear (V) (Applied Technology Council (ATC) 2009). The FEMA P695 overstrength corresponds to the yield-collapse overstrength factor +$,: - = *+$, = (. (4) If we assume the system is efficiently designed, the frame yield would be the design strength of the system. For this case we can take the design strength (V) as the yield strength (Vy). Ductility (µ, Rd) is defined as the ratio between the ultimate roof drift (δu) and the yield roof drift (δy): 23 / =  = 010+ (5) Two important yield points are important for assessing the ductility factors: the point at which first yield occurs, and the point at which the system shows the first indication of failure. The first yield point is determined by the first sign of local yielding anywhere in the SFRS. The ultimate roof drift is determined as the point where the system has had a 20% strength loss, which could be considered failure. Failure is the “near collapse” states, or the state where the system is no longer stable. Often failure is take as a strength reduction 10% from the ultimate strength (Applied Technology Council (ATC) 2009) The NBCC 2005 breaks down ductility levels into 4 simple types, Ductile (D), Moderately Ductile (MD), Limited Ductility (LD), and Conventional Construction (CC). Higher ductility levels have more requirements for detailing as well as member sizes and plastic deformation locations. Further, different SFRS types provide different ductility values (Rd) at each level; each system and ductility type specifies detailing and hinge formation requirements to achieve specific ductility and overstrength values. Table 4.1: Steel Ductility and Overstrength Information Frame Ductility Type Symbol Moment Frame Braced Frame Rd Ro Rd Ro Ductile D 5.0 1.5 4.0 1.5 Moderate Ductility MD 3.5 1.5 3.0 1.3 Limited Ductility LD 2.0 1.3 2.0 1.3 Conventional Construction CC 1.5 1.3 1.5 1.3 Systems with high ductility factors have more specific requirements including which members will yield and where yielding will occur, this generally results in rigorous detailing (Elnashai and Mwafy 2002). 4.2 Performance Design Performance design is based on assigning different performance criteria for different event probabilities. The NBCC 2005 requires design for a seismic event with a 2% in 50 year probability of occurrence (a 2500 year return period). It requires a “life safety” design and provides some guidance on interstorey drift limits. Unfortunately no guidance is provided for 24 other performance levels in the NBCC 2005. ASCE 41 Seismic Rehabilitation of Existing Buildings provides guidance for performance based design. It provides acceptance criteria for plastic deformation for different target performance levels: operational, immediate occupancy, life safety, and collapse. A variety of design seismic events are used: frequent (50% in 50 years), occasional (20% in 50 years), rare event (10% in 50 years), and the basic safety objective (BSO) similar to the NBCC 2005 (2% in 50 years). Typically the rare and the BSO events correspond to the life safety and collapse performance levels, respectively; buildings with further requirements frequently check both the frequent and occasional events against the operational and immediate occupancy performance levels. 25 Chapter 5: Methodology This study will look at steel moment frames designed for a variety of ductility levels and compare the behavior to similar frames with the addition of crossed laminated timber (CLT) infill, as a shear wall. This type of frame system is like a hybrid of an infill shear walls system and a steel plate wall system. Similar to an infill wall system, the walls will act primarily in compression although hybrid wood and steel infill wall systems would not provide the stiffness of a masonry infill wall but has the advantage of significantly less added weight to the system. The CLT infill is connected to the frame on all sides with connectors, instead of with only the confinement of the frame itself. Also, the wall is susceptible to buckling, although the buckling strength of the wall is much larger than a steel plate wall; energy is dissipated through methods other than buckling of the shear wall. Through the use of CLT shear walls we can obtain a stiffness much great than that of typical OSB or plywood shear walls. 5.1 Sample Structure The system of interest is conceived as a steel frame with a CLT shear wall placed in the plane of the moment frame. The CLT walls are connected to the steel frame with steel brackets that are nailed to the CLT shear walls and bolted to the steel frame. There would be some inevitable gap between the edge of the CLT wall and the steel members to allow for construction tolerances and potential hysteretic behavior of the brackets. The panels in this study are intended to be one solid panel within each frame bay, as oppose to multiple panels able to move relative to one another. Several models will be developed to perform a parametric study on wood and steel hybrid infill wall systems. Using a single bay frame model, the parametric studies that will be undertaken for infill system are: • the effect of frame design (design ductility levels), • infill configuration (number of infilled bays) • building height 26 A simplified floor plan and elevation are shown in Figure 5.1. The plan has 4 × 3 bays. The 2D frame elevation taken in the horizontal plan direction will be used as the base steel moment frame for analysis. The dead loads are based on concrete on metal deck for the floors for a total dead load of 4.05 kPa, and 3.4 kPa dead load for the roof. The building is assumed to have all the lateral resisting systems on the exterior walls; torsion will be ignored. For the purpose of analysis, one 2D frame will be modeled with a series of varying parameters. Figure 5.1: Base Building Floor Plan (left); Base 2D Frame Elevation (right) Initially a single bay, single storey frame was chosen for parametric study of several of the system components. The single bay, single storey frames is based on the central bay of from the building shown in Figure 5.1. The infill is placed with a gap around both sides and the top with bracket connectors on all four sides as shown in Figure 5.2. The brackets are spaced at equal spacing along each side of the panel with a maximum spacing of 800mm. 27 Figure 5.2: Single Bay, Single Storey, CLT Infilled Frame with Bracket Locations Several parameters were tested with the single bay, single storey frame to allow for a more focused approach; the parameters are summarized in Table 5.1. Table 5.1: Single Bay Parameters of Interest Parameter Range of Options Gap 0mm – 100mm gap Panel Crush Strength parallel-to-grain to perpendicular to grain compressive strength (5MPa – 30MPa) Panel Thickness 94mm – 273mm Following this, several three bay, multi-storey frame with varying infill considerations were modeled based on the building schematic shown in Figure 5.1. The gap and panel properties were set as constant. A gap of 20mm was chosen for further modeling. A panel thickness of 94mm and a crushing strength of 17.5MPa was chosen. These values are based on a three ply StructurLam panel with alternating layer orientation. The bracket spacing is similar to that described in for the single bay, single storey frame. The numbers of stories, frame design, and infill configuration are varied as described in Table 5.2. 28 Table 5.2: Multi-Storey Parameters of Interest Parameter Options Ductility Limited Ductility (LD) , Ductile (D) Storeys Three, six, and nine storeys Infilled Bays Plain frame, infilled central bay, outer bays, all bays as shown in Figure 5.3 The placement of the infill walls shown in Figure 5.3 will help determine the contribution of infill walls on the overall response of the structure. The sensitivity of the system to the area and placement of shear wall is an important factor in the implementation of the system. Figure 5.3: Wall Locations for Frame with Infill Wall Locations Different parameters will help determine sensitivity in different areas as well as determine options for ductility. The variety of SFRS types will help determine viability of different types of hybrid systems; testing each for varying ductility levels and building heights will help determine their suitability for different conditions. 5.2 Design and Modeling Computer modeling software uses finite element numerical modeling to analyze structural systems and solid models. Generally structural models are created using graphical interfaces or text files. These structural models define the material properties and the connection and boundary conditions of the system. Loads are applied to the system directly or indirectly (applied deflections or accelerations). Running the program analysis compiles all this information and returns information about the system’s response. 29 Modeling structural frames to determine the seismic response requires non-linear analysis with either static, dynamic, or response spectrum analysis. Non-linear modeling in a frame is accomplished generally by creating linear members with plastic hinges at locations of interest. Alternately, all the members could be given non-linear material properties and an appropriate mesh to determine the locations of hinging. Meshing is very important for the implementation of plastic hinges in a frame; different numerical model methods average the plastic deformation over each segment of the mesh. Further, meshing across the section of the element is important; outside flanges of an element may act plastically with the interior remaining elastic. Discrete plastic hinges avoid this by simply providing a force or moment at which plastic yielding begins and after this moment is reached a stiffness that accounts for the section shape and elasto-plastic behavior. Additionally, plastic hinges allow for faster analysis. A detailed discussion on the numerical procedure of the finite element methods used in different modeling programs is not in the scope of this paper (give some basic references, books). A brief overview of the abilities of the modeling programs used is provided. Information about the types of materials and members is given as well as a review of the analysis options offered. 5.2.1 OpenSees OpenSees is an open source, finite element software packed intended for dynamic analysis created by the Pacific Earthquake Engineering (PEER) center. The program allows for the analysis of structures and geotechnical systems under dynamic loading; it is intended to evaluate the response of structures subjected to earthquakes. It consists of modules developed to create materials and elements, assign loads or accelerations, and perform analyses. The user has significant control over the behavior of the materials and can create their own modules for material or element types if none are present that meet their needs. Further, substantial control is available to over the type of algorithms and integrators to use during analysis. The flexibility often allows for faster analysis, but can result in more numerical instability during analysis; results must be checked to avoid instability. 30 OpenSees is capable of three general types of analysis: static, transient, and variable transient. The only difference between transient loading and variable transient loading is that variable transient loading allows for a variable time step. Static loading can be either constant, for a constant load or displacement, or linear, for a linearly increasing load or displacement. Many more options exist for transient loading including multiple cyclic loading types, pulses, or time-acceleration files, such as ground motions. OpenSees is primarily a text based program using the programming language, Tcl/Tk with no graphical interface. The text based approach to this program makes it significantly less user friendly compared to commercial programs like SAP2000 or RFEM, but allow for a more parametric approach. Models can be created in a loop with one or more variables changing with each run. Graphical User Interfaces (GUIs) have been created for OpenSees, such as OpenSees Navigator, but don’t provide the same kind of parametric abilities that the text implementation allows, and often don’t allow the same level of control over output as well as the full spectrum of material and element types. 5.2.2 Frame Design and Modeling The steel frames at three, six, and nine storeys are designed to meet an equivalent static load associated with a combined RdRo of 3.0 based on testing done at FPInnovations, Rd of 2.0 and Ro of 1.5 are conservative estimates for CLT structures with nailed and screw connections (Popovski and Karacabeyli, 2011). The frames are also designed to meet the requirements for a ductile (type D) ductility moment frame and a limited ductility (type LD) as specified in the CSA S16 code (CSA-S16, 2009). The resultant frames different initial stiffness and system ductility; the stiffer frame meets the requirements of a ductile (type D) moment frame, the more flexible and weaker system meets the requirements of a limited ductility (type LD) moment frame. The members are shown in Appendix A. The system will be modeled using OpenSees. The frame members are modeled as a combination of non-linear displacement based beam columns and linear elastic beam column elements; the linear beam column elements for the center of the frame member with the non- linear beam column elements located at the ends of the member in the plastic hinge zones. The material property for the non-linear element is a bilinear material property based on the 31 moment-curvature relationships given in the ASCE41 for steel moment frames. The OpenSees model incorporated the bilinear material model based on the modified Ibarra Krawinkler Deterioration Model as shown in Figure 5.4. This model was develop to match over 300 test specimens and contains many parameters to model the moment curvature relationship as well as the strength degradation and the energy absorption of the connection types. The yield moment (My) as shown in Figure 5.4 for the steel beam and column elements is (American Society of Civil Engineers (ASCE) 2006): Beam: + = +2. (6) Column: + = 1.18+2. 41 − 66+7 ≤ +2. (7) where the effective yield strength (Fye) is determined from the material yield strength (Fy) increased by a factor of 1.1, similar to the section 27.1.7 from the Limit States Design of Steel Structures (Canadian Standards Associations (CSA) 2010). The modulus of rupture (Zx) is taken from the Handbook of Steel Construction (CISC 2010). Note that when yield capacity of the member is reduced by the ratio of the axial load (P) to the yield axial load (Py) for column elements. 32 Figure 5.4: Modified Ibarra Krawinkler Deterioration Model for BILIN Material in OPENSEES (Lignos and Krawinkler 2011) The rotation at yield (θy) is taken as shown in equations 6 and 7 (American Society of Civil Engineers (ASCE) 2006).The yield rotation is dependent on the length of the member (L) as well as the moment of inertia (Ix) and the elastic modulus of the steel (E), taken as 200,000 MPa. The yield rotation for columns is reduced as a result of axial load, similar to the moment capacity. The behavior past yield is defined based on the hinge requirements (American Society of Civil Engineers (ASCE) 2006). The plastic rotations are determined based on the class of the sections, with an upper bound set similar to a class 1 section and a lower bound set by class 3 section (Canadian Standards Associations (CSA) 2010). The definitions are shown in Table 5.3. 33 Table 5.3: Rotation Requirements for Plastic Hinges (American Society of Civil Engineers (ASCE) 2006) a b c Be a m Upper bound: 82: ≤ 52;+ and <= ≤ 418;+ 9@+ 11@+ 0.6 Lower bound: 82: ≤ 65;+ and <= ≤ 640;+ 4@+ 6@+ 0.2 C o lu m n Upper bound: 82: ≤ 52;+ and <= ≤ 300;+ 9@+ 11@+ 0.6 Lower bound: 82: ≤ 65;+ and <= ≤ 460;+ 4@+ 6@+ 0.2 5.2.3 Wall Design and Modeling CLT was implemented in the models as a linear elastic shell element. Previous analytical studies on CLT walls used rigid elements, or linear elastic shell elements. The CLT panels were based on the panels produced by StructureLam (StructurLam n.d.). These panels are produced from Spruce-Pine-Fir (SPF), often using timber affected by the pine beetle; the properties used are shown in Table 5.4. Table 5.4: CLT Material Properties (Structurlam n.d.) Material properties Design Values Elastic Modulus (E) 9500 MPa Shear Strength (Fv) 1.5 MPa Compression Strength (Fc) 30.0 MPa Compression Strength Perp-to-Grain (Fcb) 5.0MPa Poisson’s ratio 0.46 The walls are modeled as 3 ply panels with 33mm laminae, providing an overall panel thickness of 99mm. In this study, the panels were simplified to a single 99mm panel element with isotropic properties. The OpenSees model used Quad elements, elements are intended to carry stresses in the plane of the panel only, and are therefore appropriate for this type of model. The quad elements are intended for thin-plate shear element using a bilinear isoparametric formulation. It is defined by four nodes at each corner and an element thickness and a predefined material. Modeling of the walls uses shell elements with simplified strength and stiffness properties in each direction. The material applied to the quad elements was the ndMaterial 34 ElasticIsotropic. As the material name implies, the behavior of the material is both elastic and isotropic. The use of linear elastic material, or even rigid panels in previous studies justifies this simplification in the analysis. 5.2.4 Connection Design and Modeling The connection between the frame and the shear wall has two important features: the brackets connecting the shear walls to the frame, and the confinement of the wall from the frame. Testing was completed at FP Innovations on nailed steel brackets in CLT walls; the resulting hysteretic envelope is shown in Figure 5.5. Figure 5.5: Failed Bracket (left); Bracket Stress-Strain Curve (right) (Schneider 2009) The brackets were implemented in the model as twoNodeLink elements with independent behavior in each direction. The axes of the links are defined by the vector between the nodes and the order that they are defined as shown in Figure 5.6. Materials are assigned to represent the behavior in each of the directions intended to have stiffness. In this case, the bracket behavior shown is assigned in the shear direction, and a combination of the bracket -50 -40 -30 -20 -10 0 10 20 30 40 50 -40 -30 -20 -10 0 10 20 30 40 F o rc e ( K N ) Displacement (mm) Perp Para 35 and confinement behavior is assigned in the axial direction. Figure 5.6: Two Node Link Formulation (Schellenberg n.d.) The material models used to define the behavior of the brackets in the frame, the Pinching4 OpenSees material model was used in parallel with the Elastic Perfectly Plastic gap material. The analytical material model test was completed using an analytical equivalent to the CUREE testing protocols used in the experimental tests performed by Schneider (2009). The Cyclic loading profile is based on the CUREE loading profile developed for wood framed house testing in the United States. The protocol involves a series of primary and trailing cycles. The primary cycle amplitude is based on a defined percentage of the expected ultimate displacement (δu) with trailing cycles defined at 75% of its primary cycle. The expected ultimate deflection is estimated based on monotonic pushover tests. The deflection at capacity is defined as the deflection when the load drops to 80% of the ultimate load. 60% of this value is taken as the expected ultimate deflection of the system under cyclic loading; the reduction is due to the expected deterioration in strength due to cumulative damage. The sequence of primary and trailing sequences is shown in Table 5.5. 36 Table 5.5: Amplitude for Semi-static Cyclic CUREE Protocol Step No. of Primary Cycles Primary Amplitude (%CD) No. of Trailing Cycles 1 6 5% - 2 1 7.5% 6 3 1 10% 6 4 1 20% 3 5 1 30% 3 6 1 40% 2 7 1 70% 2 8 1 100% 2 9 1 120% 2 10 1 140% 2 11 1 Step 10 + 20% until failure 2 The Pinching4 material model is used to represent the bracket behavior. It is able to match the envelope of the cyclic material as well as the reloading and unloading strength and deformation, cyclic degradation of unloading and reloading stiffness and strength, and energy dissipation under cyclic loading. The Pinching4 material implementation is shown in Figure 5.7 along with the semi-static cyclic test results (Schneider 2009); the comparison shows a good match. Figure 5.7: Comparison of Test and Analytical Bracket Behavior 37 If we examine the envelope of the test data compared to that of the analytical model data we see that the two match very well as shown in Figure 5.8. The initial stiffness, and strength degradation, and damage are also a good match. Figure 5.8: Test and Analytical Bracket Behavior Comparison of Important Values The Elastic Perfectly Plastic Gap (EPPG) OpenSees material model is used to model the confinement of the panel by the surrounding steel frame. In the hybrid system proposed, the brackets do not behave alone, but interact with the frame. As the system deflects, the brackets deform, and at some point the frame may come in direct contact with the CLT panel. The deformation in the bracket after this will not remain typical. Further, the CLT panel may begin to crush if the forces are high enough at the contact point. All of this is incorporated into the behavior of the bracket using the EPPG material model. The behavior is shown in Figure 5.9. 38 Figure 5.9: Elastic Perfectly Plastic Gap Material Behavior (Schellenberg n.d.) The EPPG material model can be defined to have positive or negative stress/deformation, but not both. This is well suited to the compression only behaviour of wood crushing. There are three zones of behaviour in an EPPG material, the gap zone, the elastic zone, and the plastic zone. The gap zone has zero stiffness and is defined by the gap displacement. The elastic zone is the zone prior to yield, defined by the stiffness. The plastic zone is the zone after yield occurs and is defined by the yield force as well as the post yield stiffness as a ratio of the initial stiffness. Finally, there is also a damage parameter which allows you to define the damage as being elastic or permanent after yield has occurred. For the purpose of this study, the stiffness of the material was defined as the stiffness of the CLT panel section as defined in Table 5.4 and a gap of 20mm was assigned prior to the elastic compression response. The yield strength defined as a combination of the perpendicular and parallel crushing strength, amounting to 17.5MPa averaged over 200mm wide section of the panel. The post yield stiffness was set to 10% of the of the panel stiffness in an attempt to account for the post yield strength perpendicular and parallel to the grain. Damage was allowed to accumulate in the post-yield phase of the project due to the fully plastic nature of crushing. The effect is an increase in the size of the gap zone corresponding the displacement at the post yield load of the previous cycle. 39 The pinching material and the EPPG material are combined using a parallel formulation. In a parallel model, the strains are set to equal, and the stresses/stiffness’s are additive. In the gap zone, the only stiffness contributing to the bracket links performance is the pinching material model, and past that compression deflection, the stiffness and strength of the bracket and the panel crushing are combined. The results are shown in Figure 5.10. When under compression the stiffness of the panel dominates the behavior of the bracket. Figure 5.10: Combined Bracket Material 5.3 Loading Several loading methods are used to analyze the systems: • semi-static monotonic loading, • semi-static cyclic loading, and • transient ground motion loading. For each of these methods of analysis it is important to make sure appropriate gravity loads or masses have been applied to the model. The weight of the structure was discussed in section 5.1. Gravity loads are accumulated at the nodes joining the columns and the beams together. Subsequently a static analysis was performed on the full system at before the time was reset to zero to allow further analysis to be performed with the effect of the gravity load already in place. Each loading is further discussed below. 40 5.3.1 Semi-Static Monotonic Analysis Monotonic or Pushover analysis is when a point load, series of point loads along the height of the structure, distributed load over the height of the structure is applied. The load(s) is then increased linearly and the deflection of the structure is measured. The choice of load application shape is very important. Generally for multi-storey buildings point loads are applied at each floor/roof level. The scale of each load is important and creates what is called the load shape. There are several different load shapes frequently used in pushover analysis including: uniform distribution, inverted triangular distribution, first mode shape distribution, and code static earthquake load distribution. In some cases, the load shape is adapted as the structure moves into the plastic range and the effective period of the structure begins to change. The load shape has increasing importance for tall buildings as higher modes have a larger effect on their seismic response. For the purpose of this study, the code static equivalent load distribution described in Chapter 4 will be used. This is based on an inverted triangular distribution with additional load placed at the top of the structure in an effort to account for the typical form of the first mode shape. This simple distribution should be sufficient given the buildings on which the frames are based are a maximum of nine storeys. 5.3.2 Semi-static Cyclic Analysis Semi static cyclic tests are generally performed to help to determine the hysteretic behavior of the test subject. This can be used to determine the elastic and inelastic properties of the element or system which may be used to help determine the seismic behavior. The semi- static cyclic analysis performed is based on the ISO protocol. The cyclic displacement is based on a series of reversed cycles are determined based on the predicted ultimate displacement (01), usually obtained from a static monotonic pushover test. The peak displacement at each cycle is defined as a percentage of the predicted ultimate displacement a shown in Table 5.6. 41 Table 5.6: Amplitude for Semi-static Cyclic ISO Protocol Step Number of Cycles Amplitude of Cycle (%CD) 1 1 1.25% 2 1 2.5% 3 1 5% 4 1 7.5% 5 1 10% 6 3 20% 7 3 40% 8 3 60% 9 3 80% 10 3 100% 11 3 120% 12 3 140% For experimental tests generally the test is completed in 1 to 30 minutes depending on the displacement capacity of the test subject. For the finite element analysis performed here-in the loading will be applied in displacement steps of 0.1% of the ultimate displacement. The load distribution applied has the same form as that applied for the semi-static pushover response. 5.3.3 Ground Motion Analysis Compared to the semi-static forms of analysis, masses are applied at the nodes intersecting the columns and beams. The masses correspond to the gravity loads described in section 5.1; they are input in kg. To determine the period range over which to scale the ground motions, modal analysis was used to determine the fundamental period of each multi-storey frame. Higher modes were ignored because higher mode effects rarely have significant impact on the response of low and mid-rise buildings. The periods for each multi-storey system are shown in Figure 5.11. 42 Figure 5.11: Static Pushover Initial Periods Based on the period range determined from the static pushover analysis ground motions were chosen to match Vancouver’s 2% in 50 year hazard spectrum determined using the GSC Open file 4459 (Adams and Halchuk 2004). Ten ground motions were chosen to match Vancouver’s 2% in 50 year hazard spectrum in over the period range of 0.7 to 2.5s using the PEER ground motion database (Pacifiec Earthquake Engineering Research Center (PEER), Berkely 2011). The ground motions were then scaled linearly to match the 2%, 10%, and 50% in 50 year hazard spectrums over the range of fundamental periods in the range shown in Figure 5.12 (Adams and Halchuk 2004). 43 Figure 5.12: Scaled Ground Motion Spectrums Compared to Vancouver’s Hazard The method for linearly scaling the ground motions was to match the area beneath each ground motion response spectrum and the target hazard spectrum for the period range chosen. For each ground motion at each target spectrum a multiplier was determined to adjust the original ground motion. The scaling factors are provided in Scaled Ground Motions. 44 Chapter 6: Results For the purpose of this study the following information files were output after each analysis. • Base reactions • Total displacement at each storey • Forces in non-linear beam and column elements • Forces in bracket links • Displacement in bracket links • Stress in panel elements For each output file associated with the response of a specific element, output files were also created listing the element IDs to allow determination of the where in the model certain behaviors were occurring. 6.1 Single Bay, Single Storey, Static and Semi-Static-Cyclic Results Semi-static monotonic and cyclic analysis was performed to determine the response of the system described in Figure 5.2. It is a simple finite element model requiring limited non- linearity and is much simpler than larger multi-storey, multi-bay systems. The simplicity of this model makes it well suited for element level parametric study described in Table 5.1. The first parameter analyzed was the thickness of the CLT panel. An infill panel was implemented in a bare steel frame with the varying thicknesses commercially available from StructurLam: 99mm, 169mm, 239mm, and 309mm. These thicknesses correspond with typical three-ply, five-ply, seven-ply, and nine-ply CLT panels. An analysis was also performed on a much thinner, 10mm panel and a bare frame for the purpose of comparison; the 10mm panel is analogous to plywood. The results can be seen in Figure 6.1. 45 Figure 6.1: Effect of Infill CLT Panel From Figure 6.1 it is immediately evident that the addition of any thickness CLT panel greatly increases the stiffness and capacity of the system. Further, the initial response of the system is the same for all panels. This is a result of the gap between the panel and the frame, the additional stiffness is due to the bracket connection between the panel and the frame. At approximately 1% drift, the panel comes in contact with the frame, at which point the difference in the stiffness of the panel due to its thickness begins to have an effect. It is notable that this effect is minimal, really only noticeably different for the extremely thin panel. The panel appears to remain elastic for the commercially available panels up to 2% drift. As the panel begins to crush, the difference between the thicknesses becomes slightly more prominent between the different thickness CLT panels, but is still minimal. There is no discernible difference in ultimate strength between the different thickness CLT panels, although there is a slight increase in drift capacity of the frame as the panel thickness decreases. Comparatively, the ultimate strength is much lower for the 10mm panel. This implies that the strength of the steel frame is governing the ultimate strength of the frame for typical panels. As the panel thickness decreases enough, the ultimate strength of the system is limited by the frame. After a certain stiffness and strength, additional panel thickness does not improve the performance of the system. 46 It is important to note that the panel was only modeled as a shell element in the plane of the shell. Any potential for buckling of the panel was ignored; for thin panels buckling would likely govern the behavior of the panel instead of crushing. This would result in a significant decrease in the ultimate strength of the system for thin panels, such as the 10mm example. Another parameter analyzed is the effect of the crushing capacity of the CLT panels. The 99mm panel was implemented in a bare steel frame with varying crushing capacities in the bracket link elements. The crushing capacities were 5.0 MPa, 12.5 MPa, 17.5 MPa, 22.5 MPa, and 30.0 MPa. The 5.0 MPa capacity corresponds with the SPF perpendicular-to-grain compression capacity and the 30.0 MPa capacity corresponds SPF parallel-to-grain compression capacity. For comparison a CLT panel with infinite compression capacity was analyzed as well as a bare moment frame. The results can be seen in in Figure 6.2. Figure 6.2: Effect of Wood Crushing Strength The effect of variation of the wood crushing strength has a nominal effect on the ultimate strength of the system, but a more significant effect on the drift capacity at the ultimate strength. The behavior of the systems is identical to approximately 1% drift, which corresponds to the displacement correlated with the closing of the gap. After the gap closes the model behaviors begin to diverge for the different crushing strength until they become unstable. All the crushing strengths show similar ultimate capacities but very different drift capacity at the ultimate strength. The analysis shows that lower crushing strengths result in 47 higher drift capacities, a desirable behavior. As discussed previously, both the hysteretic behavior of the metal brackets and the crushing behavior of the panels have been incorporated into the behavior of the model bracket links. The bracket link forces with a 17.5MPa crushing strength are shown compared to the system drift in Figure 6.3. Figure 6.3: Force in Bracket vs. Overall System Drift for all Brackets Along Both Columns The bracket element numbers are referred to in the legend. The reference numbers refer to the column to which the bracket is attached (2-x for the left column, 8-x for the right column), and the bracket number from bottom to top. Negative force values represent compression. The compression forces observed in the brackets are much larger than the tension forces as is expected. When the brackets reach approximately 25 kN of compression the stiffness of the stiffness of the bracket links in compression increases considerable; The 25 kN compression force corresponds with the 20mm deflection, the specified gap size, in the brackets as shown in Figure 5.7. The compression in the brackets continues to increase approximately linearly until a system drift of approximately 1.2% is reached after which the stiffness of the bracket link model decreases considerably, due to panel crushing. This continues until the model becomes unstable at approximately 5% drift. Notably, the brackets that reach close the gap, and subsequently crush at the lowest system drifts are 2-2 and 8-6, the bottom left corner and the top right corner. This implies that the behavior of the system is that of a compression strut. 48 The final parametric study performed on a single bay infilled frame was regarding the size of the gap between the frame and panel edge. The gap can be provided to accommodate construction tolerances as well as to develop hysteretic behavior in the links; when the gap closes the stiffness of the system increases until the frame or panel yields. Pushover analyses were performed for frames with gaps ranging from 3mm to 100mm. Analyses on the bare steel moment frame as well as an infilled moment frame without a gap, and finally an infilled system without any confinement were also performed. It is important to note that an unconfined system would require the panel to be placed outside the plane of the frame which would result in torsion in the frame not considered in the analysis; this option is only provided for comparison. The results are shown in Figure 6.4. Figure 6.4: Effect of Confinement Gap on a Frame with Infill Panel All the infilled frames show the similar behavior initially and begin to increase in stiffness compared to the unconfined infilled frame at different system drifts. The system with no gap and a 3mm gap have nearly identical behavior as shown n The system without a gap increases in stiffness’ first, followed by the system with a 3mm gap, then the 10mm gap and so on. As expected, the drift at which the system stiffness increases as the size of the gap increases; more deflection is able to occur prior to the panel coming in contact with the frame for systems with larger gaps. 49 The stiffness of the systems with gaps small enough that the gap closes while the steel frame is still in the elastic range all show similar post-gap stiffness’s. The post-gap stiffness yields slightly at an approximate increase in system drift of 0.2% which corresponds to the panel increase in system drift shown in the post-gap, elastic panel zone of Figure 6.3. Finally the system continues to deflect while the panels crush; the stiffness decreases further at the drift associated with the steel frame yield. The systems where the gap closes after the steel frame begins to yield show a similar post-gap elastic panel zone, followed by a panel crushing and steel yielding zone. For large gaps such as 50mm and 100mm, a slight decrease in strength of the system was observed prior to the gap closing. To maximize both strength and drift capacity avoiding any decreases in dips in capacity, a 20mm gap shows good behavior. Along with increasing the ultimate drift capacity of the system, the gap can contribute the hysteretic behavior of the system. In a typical CLT wall system, the ductility is present in the connectors. By allowing the connectors to experience deflections, we can dissipate energy in the deformation of the nails and the brackets themselves. Unlike a plane CLT wall system, the crushing of the panels can also help dissipate the energy. A comparison of the hysteretic behavior of the plain steel moment frame, the unconfined infilled frame, and the infilled frame with a 20mm gap are shown in Figure 6.5. Figure 6.5: Cyclic and Monotonic Behavior of Single Bay Infilled Frame. Infilled Frame with 0.0mm Gap (left); Infilled Frame without Confinement (center); Bare Moment Frame (right) 50 The hysteretic behavior of the plains steel moment frame is extremely regular, as is expected from steel. There is no noticeable strength degradation due to cyclical behavior which is expected in drifts less than 4%. The unconfined infilled frame shows the effect of the brackets alone on the hysteretic behavior. The brackets do cause some strength degradation, which is expected (Schneider 2009). The hysteretic behavior also shows some stiffness degradation for the system as a whole due to the brackets, although the effect is much less severe than that is shown in the brackets alone in Figure 5.8. The hysteretic behavior of the confined infilled frame with a 20mm gap does not shows larger strength and stiffness degradation than the unconfined system due to the crushing of the panels. Overall, both infilled frame systems have effectively eliminated the pinching behavior observed in plain CLT wall systems shown in Figure 2.2, therefore significantly more energy dissipation. 6.2 Multi-Storey Static Results A monotonic pushover analysis was performed for multiple storey heights, frame ductilities, and infill configurations. Three, six, and nine storey frames with three bays were modeled as bare frames and infilled frames. Infilled frames with one, two, and three of the frame bays infilled were modeled. 99mm thick CLT panel with a crushing strength set to 17.5MPa and 20mm gaps were considered for the infilled frames. Each system was pushed to either model instability or 10% drift. The steel yield points and panel crushing points are identified on the pushover curves. The steel yield values are output from the model when the material model for the element is created for each non-linear member. The panel crush values are the same for all bracket link elements. Figure 6.6 shows the results of these pushover analyses. The pushover curves show a significant increase the strength of the system as frame bays are infilled with CLT panels, as one would expect. The drift capacity of the systems also appear to decrease as additional bays are infilled. 51 Figure 6.6: Parametric Monotonic Pushover Curves 52 The ultimate strength and stiffness relationship to the number of infilled bays is shown more clearly in Figure 6.7. Here we can see that when all frame bays are infilled, the increase in ultimate strength of the system increase in a similar pattern for both ductile (D) and limited ductility (LD) frames at all storey heights. Further, we see an increase in strength strongly correlated with the additional length of wall added. Compared to the bare steel frame, we see each infill configuration showing strengths 18%-25%, 61%-71% and 78-82% increase in strength with the addition of, respectively, single, double and all three infilled bays. These strength increases corresponds to 6m, 18m, and 24m of CLT infill, respectively, or 25%, 66%, and 100% of the frame infilled. The effect of infill CLT panels on the strength of the system is effectively linear to the percentage of the width of the frame infilled. The impact of infill CLT panels on the ultimate strength of the system is apparently independent of the number of storeys in the system or the ductility of the steel moment frame. Figure 6.7: Ultimate Strength Compared to The Number of CLT Infilled Bays The deflection at the ultimate stress is also significantly affected by the addition of CLT infill walls, although the changes in the drift behavior are more complex and are shown in Figure 6.8. Although a decrease in drift capacity at ultimate strength is expected, this is not always the case. For ductile (Type D) frames the three storey frame shows a drift capacity nearly equal to the bare frame drift capacity (7% decrease) and an increase in drift capacity is observed for six and nine storey frames with the single infilled bay CLT configuration (32% 53 increase for both). The same configuration for limited ductility frames (Type L) shows a slightly more significant decrease in drift capacity for the three storey system (12% decrease), a nearly constant drift capacity for a six storey system (4% decrease), and an increase in drift capacity for the nine storey system (15% increase). As further bays are infilled, the ultimate drift capacity begins to decrease again. For the two infilled bays CLT configuration all storey heights and frame types observed a reduction in drift capacity at ultimate. The decreases in drift capacity are somewhat less significant for the limited ductility steel frame systems (Type L) compared to the ductile steel frame system. The overall reduction shown for the three storey frame is 23% compared to 33%, for the six storey it is 21% compared to 28%, and the nine storey shows 31% compared to 28% for the limited ductility and ductile systems, respectively. Figure 6.8: Ultimate Drift Capacity Compared to the Number of CLT Infilled Bays The reduction in drift capacity at ultimate strength levels out for all systems as the infilled bay configuration moves from two to three; the reductions range from 5%-10% maximum. As larger drift capacities are preferred, for a system with minimal infilled bays, a ductile moment frame is preferred; as the number of frame bays infilled increases, there are diminishing return for the added cost of providing a ductile moment frame compared to a limited ductility frame. 54 The yield strength and drift capacity of the systems is also significantly impacted by the addition of infilled bays. This can be seen in Figure 6.9 and Figure 6.10. Figure 6.9 shows the increase in the force at which the frame begins to yield, as well as the force where the panels beginning to crush relative to the number of bays infilled. The force at which panel crushing begins increase almost linearly with the number of bays infilled with CLT panels. This behavior is independent of frame ductility and number of storeys. For both the six and nine storey frames the addition of infill CLT the panels begin to crush with a strength higher than that of the bare frame. Further, the yield strength of the steel frames increases at a greater rate than that of the CLT panels; in all cases, the CLT begins to crush prior to the steel yield. Panel crushing is a preferred type of plastic behavior over steel yield due to the ease of CLT panel replacement compared with moment frame connections. Comparatively the three storey frames show a decrease in yield strength for the limited ductility frame as well as a panel crushing strength below the bare frame yield strength. Comparatively the yield strength of the ductile moment frame increased with the additional in CLT infill walls, but the system panel crush strength is lower than the bare frame yield strength of the system. 55 Figure 6.9: Yield Strength Compared to the Number of CLT Infilled Bays 56 Figure 6.10 shows the decrease in drift capacity for the frame yield, and panel crushing relative to the number of bays infilled. The drift capacity at panel crushing is reduced from the bare frame drift capacity at first yield from 20% to 30% with the addition of one infill panel. Shifting to the two panel configuration causes a further reduction in drift capacity at panel crushing of 20%, for a total of 40-50% drift capacity reduction. The reduction in drift capacity is similar for the steel frame yield for three storey frames. The ductile frame shows notably more reduction in steel yield drift capacity with the addition of a single infilled bay, but the difference disappears as further bays are infilled. This is important because of the significant increase in strength in the system with little impact on the yield drift capacity. This is apparent for all frame types and all storey heights. The increased strength is likely due to the CLT panels engaging as a compression strut in the system, preventing pushing up the base shear needed to cause enough drift in the frames for yielding. It is also important to note that the base shears associated with the crushing of the panels begin near the point where the frames begin to yield. This is representative of the continued strength increase in the wood as it crushes. 57 Figure 6.10: Yield Drift Capacity Compared to the Number of CLT Infilled Bays 58 6.2.1 Ductility Factors The ductility and overstrength are determined using the yield and ultimate strength and drift capacities as described in section 4.1. Both the ductility and overstrength have been calculated based on the output for each multi-storey frame and infill configuration. These results are compared to the bare frame analysis results and the National Building Code of Canada design values for both a limited ductility bare frame and a ductile moment. A sample calculation for the ductility of the bare nine storey ductile frame based on the displacement and strengths at yield and ultimate as shown in Figure 6.7 to Figure 6.10. / =  = 010+ = 169.0mm 29.8mm = 5.67 Figure 6.11 summarizes the ductility(s) of the system(s) at all system heights, frame ductilities, and infill configurations. To determine the ductility the yield point must be established. Both the first yield of the steel and the crushing of the panels are considered. In Figure 6.11, the ductility and overstrength factors considering the steel yield as the yield point are denoted as “yield” and values considering the panel crush as the yield point are denoted as “crush”. Figure 6.11: Ductility Factors for Systems with Type D Frames 59 Focusing only on the ductility factors associated with steel yield, Figure 6.11 shows a decrease in the ductility of the infilled ductile steel frame (type D) system with each additional infilled bay. When one bay of CLT infill walls were added to the three and nine storey frame, the ductility reduction from the analytical bare frame ductility was 35% and 31%, respectively. The six storey frame ductilities showed an increase in ductility with the addition of the first bay of infill. This can be seen in the pushover curve shown in Figure 6.6; the single infilled bay system has a long plateau without strength degradation, whereas the bare frame shows consistent strength degradation after the yield. Similar plateau formations are seen for the nine storey frame with a less significant impact on the overall ductility. For all system heights when an additional bay of CLT infill was included, there was a further reduction 16% to 23% compared to the single infilled bay configuration. Finally, three infilled bay configuration showed very similar ductility to that of a two bay configuration, ranging from a 13% decrease in ductility to a 9% increase in ductility. Ductile frames show a general decrease in ductility with the addition of infilled CLT walls, but with each addition bay infilled, the reduction is smaller. A partially infilled ductile moment frame can show similar ductility to a fully infilled ductile moment frame. If we shift focus to the ductility factors where the yield point is determined based on the panel yield Figure 6.11 shows and increase or maintaining the ductility of the system or compared the steel yield ductility for all system heights, frame ductilities, and infill configurations. The three storey system shows ductility associated with panel crushing ranging from 43% to 54% higher for all infill configurations with the ductile (type D) frame compared to the ductility associated with panel crushing. Both the six and nine storey systems show slightly smaller ductilities, 3% and 6% lower respectively, for the single infill bay configuration for the same yield type comparison. The two and three infill bay configurations show 17% to 27% higher ductility for panel crushing instead of steel yield as the system yield point. 60 Figure 6.12: Ductility Factors for System for Type LD Frames The ductility of infilled limited ductility steel frame systems in regards to the yielding point of the steel frame (LD yield) is less affected by the addition of CLT infilled bays as shown in Figure 6.12. The variation in ductility change between system heights for each additional infilled bay is large, with no obvious trend, with some system height and infill configurations showing increases in ductility and others showing decreases in ductility. The overall change in ductility, compared to the bare frame, never exceeded 32%. Further, all system ductility, height, and infill configurations, show a ductility higher than the limited ductility value or CLT plain wall ductility value. The design value for limited ductility frames and CLT wall systems is 2.0; using the steel yield for evaluating the ductility shows an overall average ductility for 4.2 for systems with ductile frames, and 3.8 for systems with limited ductility frames. The lowest ductility value associated with steel yield was 3.1 for systems with ductile frames, and 2.9 for systems with limited ductility frames The limited ductility (type LD) frame systems show much more consistent increases in ductility for panel crushing compared to steel yield. The ductility associated with panel crushing ranges from 1% higher to 1% lower than the ductility associated with steel yield; a negligible change. The two infilled bay configuration shows ductility associated with panel crushing 18% to 27% higher than the steel yield. Finally, the three infilled bay configuration shows ductility associated with panel crushing of 23% to 49% higher than the steel yield. 61 The average ductility found where the panel crushing was used as the yield point was 5.1 for the systems with ductile steel frames and 4.4 for the systems with limited ductility steel frames; 23% and 17% higher than the respective average ductilities associated with steel yield. The minimum ductilities observed were 3.7 and 3.2 for the systems with ductile and limited ductility steel frames respectively, representing a 20% and 10% increase compared to the ductilities associated with steel yield. If the systems were designed for the panel crushing as their design elastic strength, and the frame were designed remain elastic at that value, a system ductility of 3.0 might be warranted. 6.2.2 Overstrength Factors The overstrength is calculated based on the strength of the system at ultimate compared the design strength or the strength at yield. - = *+$, = (. = 1854 KN1384 KN Figure 6.13 similarly summarizes the overstrength value(s) of the system(s) at all system heights, frame ductilities, and infill configurations. Similar to the ductility, the overstrength calculation requires the use of the yield point must be established. Both the first yield of the steel and the crushing of the panels are considered. The ductility and overstrength factors considering the steel yield as the yield point are denoted as “yield” and values considering the panel crush as the yield point are denoted as “crush”. 62 Figure 6.13: Overstrength Factors for System for Type D Frames Figure 6.14: Overstrength Factors for System for Type LD Frames shows a consistent increase in overstrength for systems with either ductile or limited ductility frames. The three storey frame shows an increase in overstrength compared the equivalent bare frame of 3% to 12% for the systems with a ductile moment frame, and 18% to 52% for the systems with a limited ductility moment frame. The six storey frame shows an increase in overstrength compared the equivalent bare moment frame ranging from 21% to 24% for ductile moment frame systems, and 1% to 9% for limited ductility systems. Finally, the nine storey frame shows an increase in overstrength compared to the bare frame 63 overstrength ranging from 5% to 10% for ductile moment frame systems, and 12% to 15%. No significant affect resulting from the number of bays infilled is apparent. Further, all system ductility, height, and infill configurations, show an overstrength near the limited ductility value or CLT plain wall overstrength value; most cases show a slightly lower overstrength. The design value for a CLT wall systems is 1.5; using the steel yield for evaluating the overstrength shows an overall average overstrength for 1.42 for systems with ductile frames, and 1.39 for systems with limited ductility frames. The lowest overstrength value associated with steel yield was 1.35 for systems with ductile frames, and 1.29 for systems with limited ductility frames. An overstrength factor of 1.3 appears to be appropriate for the overstrength associated with steel yield If we shift focus to the overstrength factors where the yield point is determined based on the panel yield Figure 6.13 shows overstrength of the system similar to or higher than the overstrength associated with steel yield for all system heights, frame ductilities, and infill configurations. Typically, the one infill bay configuration shows overstrength values associated with panel crushing similar to the steel yield overstrength values. The comparison shows slight decreases in overstrength ranging from 2% to 5% for six and nine storey systems with ductile moment frames and an increase of 40% for similar three storey systems. The comparison also shows an increase of 1% to 2% for systems with limited ductility frames at all storey heights. The two infilled bay configuration shows overstrength values associated with panel crushing significantly larger than those associated with steel yield. The comparison shows an increase ranging from 12% to 20% for all system heights with limited ductility frames, and six and nine storey systems with ductile moment frames. The three storey system with a ductile moment frame showed an overstrength value 59% higher when associated with panel crushing compared to steel. Finally, the three infilled bay configurations shows an increase ranging from 15% to 17% for systems with either limited ductility or ductile moment frames with six or nine storeys. The three storey height frames with either limited ductility of ductile moment frames show in increase in overstrength of 40% to 41% compared to the steel yield values. 64 The average overstrength found where the panel crushing was used as the yield point was 2.10 for any three storey systems, representing an increase of 30% over the average overstrength associated with steel yield. The minimum overstrength value observed for these systems was 1.67 representing effectively no increase compared to the equivalent overstrength associated with steel yield. Similarly the average overstrength observed for six and nine storey frames is 1.55 for any infill configuration and frame ductility, representing an increase of 10% over the average overstrength associated with steel yield. The minimum overstrength value observed for these systems was 1.30, representing an increase of 1% compares to the equivalent group of systems associated with steel yield. Whether the system is designed to force panel crushing before or after the steel yield, an overstrength of 1.3 appears to be warranted. This is a result in the increased yield strength of the system with the addition on infilled bays. 6.3 Seismic Response The dynamic ground motion analyses were performed on three bay models for multiple storey heights, frame ductilities, and infill configurations. Only six and nine storey frames were analyzed dynamically. Three storey frames are commonly constructed using typical stud walls and rarely need the additional capacity that steel frames and infill CLT walls would provide. Six, and nine storey frames were modeled as bare frames and infilled frames were included. Infilled frames with one, two, and three of the frame bays infilled were modeled. 99mm thick CLT panel with a crushing strength set to 17.5MPa and 20mm gaps were modeled for all the infilled systems. The ground motions were scaled to 2%, 10% and 50% in 50 years as described in section 5.3.3. The steel yield points and panel crushing points are shown on the pushover curves. The steel yield values are output from the model when the material model for the element is created for each non-linear member. To compare the behavior of the systems the drifts response of the analyses is compared for each infill configuration. To do this the inter-storey drift was averaged for the ten ground motions as demonstrated in Figure 6.15 showing the system drift at the maximum roof drift, and Figure 6.16 showing the system inter-storey drift at the maximum inter-storey drift. 65 The variation shown in the individual ground motion inter-storey drifts, especially where the distribution of the inter-storey drift over the height varies between ground motions implies some second order effect. For the purpose of this thesis, these were ignored in the scaling of the ground motions. Figure 6.15: Nine Storey Infilled Ductile Frame Building Drift at Ultimate Drift Figure 6.16: Nine Storey System with Ductile Frame Maximum Inter-storey Drift For a direction comparison between infill configurations, the average inter-storey drift is compared at each hazard level in Figure 6.17. It is clear that the addition of CLT infill panels, as expected, results in a decrease in the overall inter-storey drift of the system. This 66 is not surprising based on the overall reduction in both yield and ultimate drift of the system shown static pushover analysis in Figure 6.8 and Figure 6.10. Figure 6.17: Nine Storey Inter-Drift for all Frame-Infill Configurations It is also clear that the distribution of the inter-storey drifts over the height of the frame is heavily influenced by the addition of each CLT infilled bay. The plain frames show a more even distribution of inter-storey drift over the building height compared to the infilled systems. Further, each infilled system shows further concentration of the inter-storey drifts at the base of the structure. Effectively, with each additional infilled bay, as expected, the system behaves more like a shearwall system. This effect remains for the ground motions scaled to the lower hazard spectrums, although the change is less dramatic. Further, the infill configurations with more panels are more similar at lower hazard spectra. This can be 67 explained due to the similar elastic behavior of the systems shown in the static pushover analysis in Figure 6.6. A similar comparison in the inter-storey drift of the system is shown for the 6 storey systems in Figure 6.18. Figure 6.18: Six Storey Inter-Drift for all Frame-Infill Configurations Similarly to the 9 storey frames, the addition of CLT infilled bays causes an overall decrease in inter-storey drift of the system. Additionally, the distribution of the inter-storey drifts over the height of the frame is heavily influenced by the addition of each CLT infilled bay, similar to the nine storey systems. The overall effects are very similar, with the distribution over the height shifting that that of a typical shear wall system even more significantly. 68 Chapter 7: Conclusion Hybrid systems are commonly used in all forms of structure, with several examples existing of the hybridization of wood and steel. Design focused on the different material properties has been done in existing buildings to create efficient, sound structures. This thesis has focused on one specific form of wood and steel hybridization: CLT infill panels in a steel moment frame. 7.1 Summary of Findings The purpose of this thesis was to determine how the addition of CLT infill panels to a steel moment frame affects the behavior of the structure. The CLT panels are placed within frame bay(s), and panels are connected using typical CLT brackets bolted to the frame and nailed to the panel. The gap between the two was modeled allowing for some construction tolerances and bracket deformation; bracket deformation being the main source of ductility in a plain CLT wall system. Several parameters were analyzed, including system height, steel frame design, CLT panel thickness and strength, and placement of the CLT panel. A single bay, single storey was analyzed statically with a monotonic pushover load and a cyclic load with a range of values for the CLT panel parameters and the panel placement parameters. After the system deflects enough to close the gap between the panel and the frame, the stiffness of the system is larger as a result of the increased thickness of the panel. The ultimate strength of the system is not significantly affected, but drift at the ultimate condition increases as the panel thickness decreases. The effect on ultimate strength and the drift at ultimate is a result of the increased strength in the panel. This is seen more clearly with the analysis carrying the strength of the panel only; increasing panel strength causes a moderate increase in the strength of the system, but a significant reduction of the drift at ultimate. Finally the effect of the gap was analyzed; larger gaps result in larger increased drift at ultimate strength without showing any significant effect on the ultimate strength of the system. Several three bay, multi-storey frames were examined using both static pushover analysis and dynamic ground motion analysis. The static pushover analysis was used to determine the 69 effect of the panels on the strength and stiffness of the system, as well as to determine some preliminary values for the ductility and overstrength of the system. The dynamic analysis was used to determine the effect of the infill panels on the inter-storey drift of the system. The addition of panels was found to increase the strength and stiffness of the system almost linearly with each panel. However, the addition of panels also resulted in a decrease in the drift at ultimate, although the effect was less dramatic than the strength. A decrease in ductility with the additional panels was also noticed in the systems with significant ductility in the steel frame, but less dramatically in the limited ductility frame. Overall, the results suggest that the addition of infill bays is less beneficial in ductile moment frames; CLT infill panels are better suited to lower ductility systems. The overstrength was increased compared to the overstrength of the bare frames, but not significantly. Ultimately, a ductility factor of 3.0 is recommended for the infilled frame system, and an overstrength factor of 1.3. The addition of infill walls also reduced the overall inter-storey drift and caused the frame to behave more similarly to a shear wall system. There was some evidence of higher modes affecting the response of the taller systems; if a steel moment frame infilled with CLT walls were used as the structural seismic resisting system in increasingly taller buildings, significant attention should be paid to higher mode effects. Overall the system shows significant promise for future construction. The addition of CLT infill walls to a steel moment frame shows a significant increase in strength and stiffness of the system and reduction in inter-storey drift. Further, preliminary values indicate the panel crushing allows for ductility values higher than that of a limited ductility moment frame 70 7.2 Future Research The analytical results present several limits in the existing knowledge for some of the components as well as the overall system. To date no testing has been performed on this type of infilled wall system. I recommend the following experimental research for the development of CLT infill walls in steel moment frames as a system: 1. Experimental investigations of small pieces of CLT to determine the diagonal crushing behavior of the panel. Specifically, how the panel behaves as crushing begins, and how the panel fails 2. Experimental investigations of large pieces of CLT in plane to determine a what loads and sizes buckling becomes an issue for the panel strength 3. Experimental investigations of a single bay frame of the system monotonically and cyclically to confirm the interaction between the two systems To further develop a full design methodology for CLT infill wall systems, further analytical research is also required include the development of more detailed material models for the brackets and panel crushing based on the experimental testing programs described above. Also, a full incremental dynamic analysis as described in FEMA P695 to determine the specific ductility and overstrength factors. Finally, the concept of CLT infill walls in steel moment frames can be significantly expanded by researching other connection types. One possibility is to include multiple panels as infill within a single bay. The panels could be connected with energy dissipating connections. Strong connections with more ductility could be used between the panels and the frame, and the placement and size of the panels can be optimized to increase the efficiency of the system. 71 References Adams, J., and S. Halchuk. 2004. Fourth Generation Seismic Hazard Maops of Canada: Values for over 650 Canadian Localities Intended for the 2005 National Building Code of Canadaa. Geological Survey of Canada, Ottawa: National Eartquake Hazards Program. American Society of Civil Engineers (ASCE). 2006. Seismic Rehabilitation of Existing Buildings, ASCE 41. 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Thesis (MASc), Vancouver, Canada: University of British Columbia. Kodor, V.R., M.A. Erki, and J. Quenneville. 1995. "Seismic Design and Analysis of Masonry Infilled Frames." Canadian Journal of Civil Engineering Vol. 22 (No.3) 576-587. Koshihara, M., Isoda H., and S. Yusa. 2005. "The Design and Installation of Five-Storey New Timber Building in Japan." Summary of Technical Papers of Annual Meeting. Architectural Institute of Japan. 201-206. Krawinkler, H., F. Parisi, L. Ibarra, A. Ayoub, and R. Medina. n.d. Development of a Testing Protocol for Woodframe Structures. CUREE Publication No. W-02, Consortium of Universities for Research in Earthquake Engineering (CUREE). Krisciunas, R. 1996. "Ontario's experience with Composite Wood/Steel Bridges." National Conference on Wood Tansportation Structures. Madison, US: US Department of Agriculture, Forest Sevice, Forest, Products Lab. Lignos, D., and H. Krawinkler. 2011. 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Candian Journal of Civil Engineering, Vol. 36 (No.2) 280-294. 75 Appendices Appendix A Frame Design Figure A-1: Nine storey ductile steel moment frame 76 Figure A-2: Nine storey limited ductility steel moment frame 77 Figure A-3: Six storey ductile steel moment frame 78 Figure A-4: Six storey limited ductility steel moment frame 79 Figure A-5: Three storey ductile steel moment frame Figure A-6: Three storey limited ductility moment frame 80 Appendix B Scaled Ground Motions B.1 Ground Motions NGA # Event Year Station Mag Quake Type Rjb (km) Rrup (km) PGA (g) 174 Imperial Valley-06 1979 El Centro Array #11 6.5 Strike- Slip 12.4 12.4 0.37 185 Imperial Valley-06 1979 Holtville Post Office 6.5 Strike- Slip 5.5 7.7 0.26 316 West- morland 1981 Parachute Test Site 5.9 Strike- Slip 16.5 16.7 0.17 738 Loma Prieta 1989 Alameda Naval Air Stn Hanger 6.9 Reverse- Oblique 70.9 71 0.22 949 Northridge- 01 1994 Arleta - Nordhoff Fire Sta 6.7 Reverse 3.3 8.7 0.24 1077 Northridge- 01 1994 Santa Monica City Hall 6.7 Reverse 17.3 26.4 0.49 1116 Kobe, Japan 1995 Shin-Osaka 6.9 Strike- Slip 19.1 19.1 0.19 1187 Chi-Chi, Taiwan 1999 CHY015 7.6 Reverse- Oblique 38.1 38.1 0.17 1317 Chi-Chi, Taiwan 1999 ILA013 7.6 Reverse- Oblique 81.7 84.1 0.12 1489 Chi-Chi, Taiwan 1999 TCU049 7.6 Reverse- Oblique 3.8 3.8 0.28 B.2 Scaling Factors Ground Motion 50% in 50 years 10% in 50 years 2% in 50 years NGA_174 0.282 0.732 1.419 NGA_185 0.225 0.586 1.135 NGA_316 0.218 0.567 1.101 NGA_738 0.208 0.540 1.047 NGA_949 0.258 0.669 1.298 NGA_1077 0.203 0.528 1.025 NGA_1116 0.244 0.633 1.228 NGA_1187 0.257 0.668 1.295 NGA_1317 0.207 0.538 1.044 NGA_1489 0.220 0.571 1.108
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Clt infill panels in steel moment resisting frames as a hybrid seismic force resisting system
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