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Foundation Design of a Shoppers Drug Mart in Squamish, B.C. To, Martin Ho-Nang 2009-04-03

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    FOUNDATION DESIGN OF A SHOPPERS DRUG MART  IN SQUAMISH, B.C.    by  Martin Ho-Nang To      A THESIS SUBMITTED IN PARTIAL FULFILMENT OF  THE REQUIREMENTS FOR THE DEGREE OF  BACHELOR OF APPLIED SCIENCE    in  GEOLOGICAL ENGINEERING Faculty of Applied Science Geological Engineering Program           THE UNIVERSITY OF BRITISH COLUMBIA APRIL, 2009  iiABSTRACT A foundation design is needed for a proposed Shoppers Drug Mart, located in the Chieftain Shopping Centre in downtown Squamish, British Columbia. The purpose of this study is to investigate the soil conditions at the proposed building site, complete settlement and liquefaction analysis, and provide recommendations for earthwork and foundation design.   Firstly, a comprehensive geotechnical site investigation will need be performed, including two Cone Penetration Test (CPT) boreholes and four auger drillholes. Thereafter, using the data obtained from these tests, as well as performing lab experiments with the soils, soil properties can be determined for the soil stratigraphy of the proposed site. Using these soil properties, a foundation type for the building can then be considered. However, one major issue to consider prior to designing for the foundation is since the proposed site is located in a seismically active region, seismic design considerations will need to be taken in the design. The following report will present the findings of the soils investigation and our recommendation for geotechnical aspects of the project.  The pre-existing building for the proposed development was needed to be demolished prior to constructing the new one. The new proposed development will include the construction of a single storey commercial building about 22 feet high with a maximum surface area of 122 feet by 158 feet.   iiiACKNOWLEDGEMENTS    I would like to thank the following individuals for their contributions and assistance for this report.   • CENTENNIAL GEOTECHNICAL ENGINEERS LTD.: o Mr. Louis Lui o Mr. Nouver Cheung    • UNIVERSITY OF BRITISH COLUMBIA: o Dr. Ulrich Mayer o Dr. John A. Howie o Dr. Dharma Wijewickreme    TABLE OF CONTENTS  1.0  INTRODUCTION ......................................................................................... 1 1.1  Location ..................................................................................................... 1 1.2  Proposed Development .............................................................................. 2 2.0  SITE INVESTIGATION............................................................................... 4 2.1 Site Physiography ....................................................................................... 4 2.1.1 Topography............................................................................................ 4 2.1.2 Vegetation.............................................................................................. 4 2.1.3 Land Use ................................................................................................ 5 2.1.4 Climate................................................................................................... 6 2.1.5 Drainage................................................................................................. 7 2.2 Bedrock Geology ........................................................................................ 8 2.2.1 Formation of Basement Rocks............................................................... 9 2.3 Groundwater Conditions............................................................................. 9 2.4 Geological Hazards................................................................................... 10 2.4.1 Risk of Volcanoes and Resulting Landslides....................................... 10 2.4.2 Earthquakes.......................................................................................... 12 3.0  METHODS OF INVESTIGATION........................................................... 15 3.1 Subsurface Investigation........................................................................... 15 3.1.1 Auger.................................................................................................... 15 3.1.2 Dynamic Cone Penetration Tests......................................................... 16 3.1.3 Cone Penetration Tests ........................................................................ 16 3.2 Lab Testing ............................................................................................... 20 4.0  SEISMIC DESIGN CONSIDERATIONS................................................. 21 4.1  Seed’s Simplified Analysis...................................................................... 21 5.0  FOUNDATION DESIGN OPTIONS......................................................... 28 5.1  Site Preparation........................................................................................ 28 5.1.1 Preloading ............................................................................................ 28 5.1.2  Dynamic Compaction ......................................................................... 31 5.1.3  Vibro-Compaction Processes.............................................................. 32 5.2  Shallow Foundations................................................................................ 33 5.2.1 Strip/Spread Footings........................................................................... 33 5.2.2  Mat Foundation................................................................................... 36 5.3  Deep Foundations .................................................................................... 40 5.3.1  Timber Piles........................................................................................ 40 5.3.2  Steel Piles............................................................................................ 41 5.3.3  Concrete Piles ..................................................................................... 44 6.0  THE CONSTRUCTION PROCESS.......................................................... 48 6.1  Vibration Monitoring............................................................................... 48 6.2  PDA Testing and Checking Integrity of Piles.......................................... 51 7.0  RESULTS AND CONCLUSION ............................................................... 53 LIST OF REFERENCES................................................................................... 55 APPENDICES A. Site Map B. Regional Geology Map C. Aerial Photographs D. Auger Drill Data E. CPT Data F. Cross Section G. Results and Calculations H. Site Photographs   viLIST OF FIGURES AND TABLES  Figure 1:  Location map of the proposed site Figure 2:  The pre-existing building site Figure 3: Illustration of the subducting plates Figure 4:  Seed’s simplified liquefaction assessment with Corrected CPT Tip Resistance vs. CSR Figure 5: Seed's simplified liquefaction assessment using (N1)60 values vs CSR Figure 6:  Illustration of the number of retaining wall blocks required for the site Figure 7:  Peak Particle Velocity relative to the distance away from the source Figure 8:  Distance vs Peak Particle Velocity values relative to damage threshold values   Table 1: Soil stratigraphy determined by means of Auger Drilling Table 2:  Data from CPT#1 and the unit weights for the soil layers Table 3:  Data from CPT#2 and the unit weights for the soil layers Table 4:  Liquefaction analysis of CPT#1 (Tip Resistance vs. CSR) Table 5:  Liquefaction analysis of CPT#2 (Tip Resistance vs. CSR) Table 6:  Liquefaction of CPT #1 ((N1)60 values vs CSR) Table 7:  Liquefaction of CPT #2 ((N1)60 values vs CSR) Table 8:  Ultimate settlements from the two CPT boreholes Table 9:  Number of retaining wall blocks required, without vertical drains Table 10:  Advantages and disadvantages of preloading Table 11:  Footing dimensions and capacities from CPT#1 Table 12:  Footing dimensions and capacities from CPT#2 Table 13:  Footing load and settlement values for CPT #1 (Classical Method) Table 14:  Footing load and settlement values for CPT #2 (Classical Method)  viiTable 15:  Typical subgrade reaction values Table 16:  Mats foundation results from CPT #1 Table 17:  Mats foundation results from CPT #2 Table 18:  Calculated bearing capacity values for mats foundation Table 19:  Calculated settlement values for mats foundation Table 20:  Pile capacity values from CPT#1 for steel piles Table 21:  Pile capacity values from CPT#2 for steel piles Table 22:  Pile capacity values from CPT#1 for concrete piles Table 23:  Pile capacity values from CPT#2 for concrete piles Table 24:  Vibration monitoring data Table 25:  Vibration monitoring close to the source Table 26: Final tested pile capacity values    11.0  INTRODUCTION 1.1  Location  A Shoppers Drug Mart is proposed to be built in Squamish, British Columbia, located approximately 70 kilometres from downtown Vancouver. The proposed site has the co-ordinates of N49° 42' 6.20", W123° 9' 10.80". Located in the Squamish Valley region, the proposed development includes a maximum plan dimension of about 22 feet in height and 122 feet by 158 feet in surface area. It will be constructed at the southwest corner of the Chieftain Shopping Centre, located in downtown Squamish at the corner of Pemberton Avenue and 3rd Avenue.   Figure 1: Location map of the proposed site. N  21.2  Proposed Development In order to construct the proposed development, a pre-existing building would need to be demolished. Surrounded by asphalt-paved parking lots at the south and east corners of the mall, the existing and proposed floor grade of the building will be at Elev. 2.86m (geodetic datum).  The ground surface of the pre-existing building slopes gently down from east to west, from approximately Elev. 2.5m to Elev. 1.8m respectively.   Figure 2:  The pre-existing building site. (Taken March 7, 2007)  Squamish is located on the Coast Belt and is surrounded by volcanic structures, thus the region is at a high risk of seismic activity. Therefore, when designing for the building, seismic design considerations will need to be taken, such as ground motion analysis and liquefaction assessment. In addition, after a site investigation is performed, a design for the foundation type that will keep the foundation intact in the event of a large earthquake will be needed for the building.    3After the building is completed, necessary testing procedures such as vibration monitoring from the construction of the foundation, pile driving analysis, and pile integrity testing will be need to be performed. This is to ensure that the piles are intact and to obtain the real bearing capacity values of the piles.   42.0  SITE INVESTIGATION 2.1 Site Physiography  2.1.1 Topography Since the township of Squamish area is a flat valley floodplain, the elevation remains fairly constant throughout the region, as a result of large scale sediment deposition in the post-glacial period by the rivers and streams. (Squamish District Council. p.12-13) However, after the glacial period, when the glaciers retreated and created the present Squamish Valley, it deposited alluvial deposits from 300 to 400 feet above the general elevation of the valley floor. To the southeast of Squamish is a 600 metres high mountain named Stawamus Chief (otherwise known as the Squamish Chief).  2.1.2 Vegetation The vegetation in the Squamish region is mostly occupied by a mixed forest of coniferous and deciduous trees. Since the growth rate is high, especially for deciduous trees, they occupy continuously throughout the area. However, before the area was inhabited, the river floodplain was occupied by mostly cedar and spruce trees, in which some of the large stumps still remain standing to this date. (Stathers, 1955, p.19-21) Regardless, these trees are gradually being replaced by deciduous trees, due to extensive logging in the area.     52.1.3 Land Use Within the lower Squamish valley, land usage is divided into five major categories: 1.) area used for buildings, townsites, and industry 2.) cultivated farm land 3.) farm land abandoned or reverted to pasture 4.) waste land – sand and gravel bars, tidal flats, and natural meadow 5.) forest – virgin forest, second growth, or recently logged (Reference – Stathers, 1955, p.60) However, since the mountains and surrounding rivers restrict land usage only to the valley itself, the development of Squamish has followed a linear pattern.  The township of Squamish has six categories of land use: downtown commercial, mixed commercial residential, tourist recreation commercial, highway commercial, residential, and industrial. (Squamish, 1989, p.34-36)  Downtown commercial is intended for the development of retail, office, personal services, institutional, entertainment, and government services. (Squamish, 1989 p.34) Mixed commercial and residential is the combination of the two, tourist recreation commercial is intended to attract recreational-tourist opportunities, and highway commercial is land for commercial uses along the side of the highway. (Squamish, 1989 p.34-36) However, with the increasing population in this  6area, more housing developments are being built north of downtown Squamish, into areas such as Brackendale and Southridge.   2.1.4 Climate Due to the high surrounding mountains and the sea, the Squamish valley receives approximately 60 to 70 inches of precipitation per year. During the summer, humidity is high on average, approximately in the range of 75 to 87 percent, and daytime temperatures typically ranging from 13 to 18ºC. However, no weather records were kept for the valley, therefore all values presented will only be an estimation. In addition, the values from the Squamish valley region would have little difference to those obtained from weather stations in the Vancouver area.  Snowfall typically occurs near the end of November or the beginning of December, and ends by the end of March or beginning of April. (Stathers, 1955, p.24) During the day, wind from the Howe Sound direction blows into the valley during the day, whereas at night the wind direction is reversed. This pattern of wind allows a fairly constant temperature during the summer, and permits warm daytime temperatures during the winter. However, north winds can occur in both the summer and winter, creating high temperatures and low humidity during the summer and low temperatures and heavy snowfall during the winter. (Stathers, 1955, p.25) These north winds during the winter time can reach 50 km/h in gusts, bringing along an out-flow of cold air from interior British Columbia and spilling over the mountainous terrain of Garibaldi  7Park. This will lead to very cold temperatures; however, such occurrences are not frequent, only happening once or twice each year. (Stathers, 1955, p.26)  2.1.5 Drainage  Most of the drainage in the Squamish valley area is provided by the Squamish River, as it is situated in the middle of the valley, and drained through the Squamish River delta, directly into the Howe Sound. The Squamish River receives large amounts of runoff, since the Elaho, Cheakamus, and Manquam Rivers all drain into this river, before entering into sea. (Stathers, 1955, p.30) In addition, the Stawamus River, which is situated on the eastern side of the valley, empties into Howe Sound as well.   Flooding in the Squamish valley area have been a continuous problem, since rapid changes in weather often occur in this area, especially in the month of October since the previous winter’s snow has disappeared from the surrounding mountains and added high temperatures and heavy precipitation to the mountains will drain the water rapidly. (Stathers, 1955, p.42) In addition, with the combination of high tides and strong winds, overflowing of the rivers will often occur, especially in the Mamquam River. Roads and bridges will be washed out, and houses will be half-filled with water. Therefore, in downtown Squamish, basements are not allowed in residential units, as a result of poor drainage in that area. Consequently, dykes were built along the rivers surrounding the Squamish region.    82.2 Bedrock Geology Through research papers and studies done by other geoscientists, it was found that the rock structure of the Howe Sound region, which includes Squamish, were formed mostly by volcanoes. It is part of the formation of the Coast Belt, which contains the Coast Mountains. This mountain range is divided into two segments: Southwestern Coast Mountains, and Southeastern Coast Mountains. Our area of interest is the Southwestern Coast Mountains, which dominantly consists of quartz diorite and granodiorite. (Monger, 1994, p.11) These formations were formed as a product of four distinct geological episodes: islands were first formed in the region approximately more than 140 million years ago; numerous granitic plutons intrusions on the islands between 140 and 90 million years ago; followed by uplifted rocks from 90 to 20 million years ago. Many volcanic activities forced the dykes and sills into the sedimentary rocks, constructing nearby mountain ranges such as Mount Baker and Mount Garibaldi. (Armstrong, 1990, p.42) Within this mountain range rests the District of Squamish, which contains the key attraction in this region, the Squamish Chief. The mountain contains a large mass of granodiorite, which is comprised of mostly early Cretaceous medium-grained granodiorite, in which the rock crystallized approximately 100 million years ago. However, even though these rocks were metamorphosed, they are unsheared, proven by the angular fragments contained inside the rock. These rocks are sparsely jointed, therefore proven to be resistant to erosion. (Mathews and Monger, p.162-163)  92.2.1 Formation of Basement Rocks  Approximately 167 to 91 million years ago, during the Middle Jurassic to mid-Cretaceous era, these quartz diorite, granodiorite, and minor diorite, with minor septa and fault slices of the Triassic and Jurassic strata belonging magmatic art built on the eastern parts of the Wrangellia and Harrison terranes and the overlapping Lower Cretaceous Gambier Group were formed. (Monger, 1994, p.11) The youngest plutons, located in the southwestern Coast Belt, ranges from the age of 110 to 91 million years (Ma) and they overlap in age with the oldest plutons, which are approximately 103 Ma.  (Monger, 1994, p.12) In the southwestern Coast Mountains, land is typically deformed along a discrete, contractional shear zones which are north-northwest trending, and dominantly west-southwest-vergent. Rocks between shear zones can span greater than 10 kilometres wide may be little deformed, and as young as mid-Cretaceous (96 Ma). (Monger, 1994, p.13)  2.3 Groundwater Conditions The groundwater flow in the downtown Squamish region was found to be comparatively non-existent, because of the flat gradient of the Squamish valley and throughout the downtown core. However, surrounding the valley are many groundwater recharge zones, which includes: lower Cheakamus Valley, Cheekeye Fan, Squamish Valley floodplain, lower Mamquam Valley, and lower Stawamus Valley. There are aquifers that surround these groundwater recharge zones, but none that surrounds our site of interest. (Squamish (B.C.) District Council, p.29)  10From the auger drill holes, found in Appendix D, the location of the water table was noted in auger holes A1, A2, and A3. As well, from the CPT logs, the water table was found to be where positive pore water pressure was. The water table was found to be approximately 2 metres in depth.  2.4 Geological Hazards The Squamish region is surrounded by potentially volcanically active mountains and the risk of melting glaciers. In addition, it is located in a seismically active zone and some parts are located in a delta or alluvial fan. With these conditions and landscape, the potential for hazards is high.   2.4.1 Risk of Volcanoes and Resulting Landslides Mount Garibaldi, one of the three major volcanic complexes of the Quaternary Cascade magmatic arc, is part of the Garibaldi volcanic belt (GVB) and the closest and largest risk to Squamish. (Monger, 1994, p.232) If a volcano were to erupt on this mountain, it would pose a serious threat to those living in the Squamish and Whistler-Pemberton area. Ash columns can rise to a few hundred metres high, and could affect the regional area’s air quality and air traffic. In addition, the flow of the lava can cause slope instability, and destroy homes, roads, and river flows in its path. However this risk is low to moderate since lava tends not to travel too far from its source. Additionally, with lava melting the ice and the resulting ashfall, or tephra, it could contaminate the water supply for the regional area since the catchment area for the regional watershed is downwind from Mount Garibaldi. (GSC, 2005)   11 The Garibaldi volcano is situated on crystalline basement rock from the Coast Mountains. (Monger, 1994, p.239) Andesitic and dacite lava flows and pyroclastic rocks outlines the ridges of Mount Garibaldi, which was formed near the end of the ice age. The resulting debris lava flow (lahars) from the volcano flowed off the sides of the mountain, eroding the ice off the side of the peak, leaving behind a steep drop-off on the southwest-facing cliffs, and causing the resulting debris avalanche materials consisting of dacitic lavas and tuff-breccias to be accumulated and deposited into the Cheekye Fan, which consists of a upper kame terrace and a lower alluvial fan. (Monger, 1994, p.239)    As well, the resulting lava from the volcano can cause the ice and snow on the mountains in that area to melt very quickly. Most of these materials from Mount Garibaldi will likely end up in the Cheekye River, down the Cheekye valley into the Cheekye Fan, Cheakamus River, or Rubble Creek. However, these events can cause potential catastrophic floods and landslides to areas below these mountains, even affecting areas such as the downtown Squamish and other neighbouring communities such as Brackendale. As well, landslides can occur from a source named The Barrier, which is a steep rock face formed by successive failures of the margin of the Clinker Peak lava flow. (Monger, 1994, p.267) Clinker Peak is in close approximation to Mount Garibaldi, but the rock avalanche materials follow a different path, as most of the debris ended up at a large fan at the mouth of Rubble Creek sometimes down to the Cheakamus  12Valley. (Monger, 1994, p.267-269) Consequently, landslides can affect Highway 99, which is the main artery that connects Whistler and Squamish with Vancouver. In addition, landslides can completely destroy communities and deposit unwanted sediments into the river. (GSC, 2005) 2.4.2 Earthquakes In the southwestern portion of British Columbia, the chance of earthquakes is very high, as this region is a seismically active zone. There are three distinct source regions for earthquakes in this region: earthquakes within the continental crust, deeper earthquakes from subducted oceanic plates, and earthquakes on the subduction boundary between lithospheric plates. (Monger, 1994, p.221) Nonetheless, subduction earthquakes are among the world’s largest earthquakes, therefore buildings and structures in this region will have to follow the National Building Code of Canada (NBCC) for seismic hazards caused by horizontal ground shaking.   Earthquakes within the continental crust in the southwestern British Columbia region typically are small earthquakes. However, these earthquakes usually happen at a considerable depth within the crust, at approximately 20km depth, therefore aftershocks are much less than typical California earthquakes, which most earthquakes occur at the top 10km of the crust. (Monger, 1994, p.222) These small crustal earthquakes are a mixture of strike-slip and thrust event with a dominant north-northwest orientation of the principal stress axes suggesting north-northwest compression (Monger, 1994, p.224)   13 Subcrustal earthquakes in this region refers to the subducting Juan de Fuca Plate, in which this plate is very thin and shallow in depth (maximum 10km depth), therefore aftershocks are rare. (Monger, 1994, p.225) Since the plate is very brittle, the maximum magnitude an earthquake can create is 7. There are 2 locations in which the plate can dip under: west coast of Vancouver Island, and below the Strait of Georgia and Puget Sound. (Monger, 1994, p.225) However, these earthquakes can still cause considerable damage, as demonstrated in earthquakes occurring in 1949 and 1965 at the south end of Puget Sound, with magnitudes in the range of 5 to 6. (Monger, 1994, p.225) Subduction earthquakes are the strongest and most devastating, providing magnitudes of 8 or greater. In the event of an earthquake, the most intense area will be the subduction boundary between the Juan de Fuca Plate and the North American Plate, as shown in Figure 6.  This kind of earthquake is long in duration of strong shaking, mainly associated with large rupture surfaces. In addition, the area of shaking is large compared to other forms of earthquake; therefore it can affect certain types of major structures, such as tall and large structures. The heavy weight of these buildings will increase the liquefaction potential of saturated sands. (Monger, 1994, p.228) These subducting plates are responsible for the formation of the Cascade Range as well as the Pacific Range. However, this type of earthquake is rare, with the last major earthquake of this type happening approximately in the year 1700. (Monger, 1994, p.227)  14 Figure 3: Illustration of the subducting plates <http://gsc.nrcan.gc.ca/geodyn/images/mega12.jpg>     153.0  METHODS OF INVESTIGATION Prior to the construction of the building, an investigation of the soil stratigraphy of the site was needed. Several testing programs were implemented for this investigation, which included subsurface investigations using an auger drill and performing cone penetration tests. In addition, lab testing which included obtaining the in-situ moisture content of the soil was also performed. Locations of the boreholes are indicated in Appendix A.   3.1 Subsurface Investigation Two different soils investigation programs were implemented for the site: an auger drill, and a Cone Penetration Test (CPT). The auger drill went to a depth of 20 to 40 feet, whereas the CPT was done to a depth of 100 feet. The auger and CPT logs are located in Appendix D and E respectively.   3.1.1 Auger Through four different auger drill holes, located around the perimeter of the site of the proposed building, a conclusion of the soil stratigraphy of the site can be determined, as follows: 0 – 4 feet Dense Fill 4 – 8 feet Clayey, Low-Plasticity Silt 8 – 16 feet Fine-Grained, Low-Plasticity Silt 16 – 40 feet Clean, Fine to Medium to Coarse-Grained Sand Table 1: Soil Stratigraphy determined by means of Auger Drilling The dense fill was found to be mainly tan brown, typically fine to medium-grained silty sand with occasional gravel. The layer below is the clayey, low-plasticity silt which contains some organics. This layer is very soft, and will potentially liquefy upon earthquake loading, depending on  16its plasticity index and liquid limit. This may cause a foundation such as a footing foundation to experience a punching shear failure. The layer beneath is similar to the one above it, as the soil has low plasticity and contains some organics as well. However, this layer is grey in colour and contains some very fine-grained sand, which increases the chance of liquefaction upon an earthquake because of the presence of sand. Finally, the last layer is the sand, which is compact, clean, grey in colour, and is typically fine to medium-grained which coarsens to coarse-grained with depth.  3.1.2 Dynamic Cone Penetration Tests Upon completion of the auger drill hole, typically a dynamic cone penetration (DCP) test followed and is part of the auger drilling. The DCP tests were performed on auger holes A1 and A4. This test allows the soil stratigraphy logger to confirm the soil properties that has been logged, as well as obtaining a rough estimation of the densities and N60 values of the different soil layers. In general, the shallow fill layer exhibited higher blow counts per foot, ranging from 30 to 60 blows per foot, whereas the deeper silt and sand layers lingered around the 20 blows per foot range. This confirms the fact that the silt and sand layers are not dense, therefore increases the chance of liquefaction upon an earthquake.   3.1.3 Cone Penetration Tests Two CPT boreholes were drilled on this site, one located in the northwestern corner and the other located in the southwestern corner of  17the demolished building site. Using a drill bit that has a tip area of 10 cm² and a sleeve area of 150 cm², both drillhole locations exhibited similar soil characteristics, while comparing the parameters of qc (tip-bearing resistance), fs (sleeve friction), u2 (pore pressure), and N60 (SPT N-values) to depth.  Overall, many of the layers up to 30 metres (100 feet) in depth exhibited a range in tip-bearing resistance value of 5 to 15 MPa, which would tell us that the relative density of the soil is low. However, there are two layers that exhibit particular high qc values: the sand to silty sand layer at approximately 5 to 8 metres (16 to 26 feet) with qc values of up to 20 MPa for CPT#1 and 50 MPa for CPT#2; and the sand layer at approximately 25 to 30 metres (82 to 100 feet) with qc values of up to 20 MPa for both CPT#1 and CPT#2.  While comparing the sleeve friction (fs) values, it is found that the values peak at the same places as the tip-bearing resistance values. Typically, the layers display a sleeve friction value ranging from 20 to 100 kPa. However, at the same range of depths as where high qc values were noticed, fs values as high as 600 kPa occurred in the sand to silty sand layer and 125 kPa in the sand layer.   During the observation of the pore pressure measurements, it is determined that the water table is located at 2 metres (6.5 feet). Pore pressure starts off as negative from the ground surface to the water table, and as depth increases below the water, pore pressure measurements  18gradually increased as well. However, at approximately 5 to 8 metres (16 to 26 feet), once again, exceptionally high pore pressure values of up to 35 MPa were observed for both CPT boreholes. This behaviour is caused by the thin lenses of silt contained within the sand layers, causing the building up and subsequent dissipation of pore pressure.  While performing the CPT, N60 values were obtained as well. The N60 values gives the blow count per every penetrated foot, therefore allowing the estimation of relative density of the soil layers. A high N60 value will indicate that the soil is dense whereas a low N60 value means the soil is loose. For our two CPT test boreholes, it was found that an N60 value of 70 was achieved at 3 metres (10 feet). In addition, at 6 metres (20 feet), the N60 value was 120. At these two locations, the soil has very high relative density, and proved through auger drilling that at those depths is the dense silty sand layer. Other than at these two depths, the N60 values generally hover around 10 to 20, with the exception at 26 metres (85 feet) for CPT#1 and 28 metres (92 feet) for CPT #2, where the N60 value was found to be 40, which was determined to be a silty sand layer. Overall, the soil conditions at the site to a depth of 30 metres are generally relatively loose.  Before performing analysis on the soil layers, soil classification and unit weights of the soils would need to be first obtained. Below are tables illustrating the soil stratigraphy and their respective unit weights at  19the given depths for the site, using data obtained from the two CPT boreholes: Depth (m)    Unit Weight  ThicknessFrom To Layer (kN/m³) (m) 0.00 1.00 Sensitive Fine Grained 17.5 1.00 1.00 3.75 Clay 18.0 2.75 3.75 5.25 Sand to Silty Sand 19.0 1.50 5.25 7.00 Sand 19.5 1.75 7.00 13.00 Sand to Silty Sand 19.0 6.00 13.00 14.75 Silty Sand to Sandy Silt 18.5 1.75 14.75 19.00 Sand to Silty Sand 19.0 4.25 19.00 23.00 Sand 19.5 4.00 23.00 23.75 Sand to Silty Sand 19.0 0.75 23.75 27.75 Sand 19.5 4.00 27.75 29.25 Sand to Silty Sand 19.0 1.50 29.25 30.25 Sand 19.5 1.00 30.25 32.25 Sand to Silty Sand 19.0 2.00    Table 2: Data from CPT#1 and the unit weights for the soil layers  Depth    Unit Weight  ThicknessFrom To (kN/m³) (kN/m³) (m) 0.00 2.00 Sensitive Fine Grained (Dry) 17.5 2.00 2.00 3.75 Sensitive Fine Grained (Wet) 18.5 1.75 3.75 8.75 Sand to Silty Sand 19.5 5.00 8.75 13.00 Silty Sand to Sandy Silt 18.5 4.25 13.00 13.50 Sand to Silty Sand 19.0 0.50 13.50 14.75 Silty Sand to Sandy Silt 18.5 1.25 14.75 16.00 Sand to Silty Sand 19.0 1.25 16.00 22.00 Silty Sand to Sandy Silt 18.5 6.00 22.00 24.75 Sand to Silty Sand 19.0 2.75 24.75 25.50 Sand 19.5 0.75 25.50 28.25 Sand to Silty Sand 19.0 2.75 28.25 30.25 Sand 19.0 2.00    Table 3: Data from CPT#2 and the unit weights for the soil layers  203.2 Lab Testing After the soil samples were obtained during the drilling process, they were taken back to the laboratory for further testing. The testing program included obtaining the in-situ moisture content of the soil. The moisture content of the soil can give us a general idea of the strength of the soils as well as acquiring the parameters of the soil properties.  Through the soils investigation, it was found that the fill had relatively low moisture content values, typically ranging from 5% to 10%. However, the layer beneath that, the silt layer, had very high moisture content values. Since the silt contained some organics which were loose, its moisture content values varied from 33 % to 62%. Conversely, at deeper depths, the silt layer decreases in organics content and becomes firmer at the same time, with its moisture content decreasing to a range of 34% to 52%. Both of these silt layers are very much susceptible to liquefaction, since the density of the soil is low and the moisture content remains very high. Finally, the sand layer from the depth of 16 to 40 feet has typical moisture content values of 15% to 28%. This layer generally contains clean and compact, fine to medium-grained sand with occasional gravel. In addition, it possesses moderately high relative densities (approximately Dr of 70%), therefore this layer should be relatively unsusceptible to liquefaction.    214.0  SEISMIC DESIGN CONSIDERATIONS  The Shoppers Drug Mart located in the Squamish district is located in a seismically active area, therefore earthquake design considerations will need to be considered in the design of the structure. Several factors are taken into account in the design process: • The intensity and magnitude of the earthquake • The depth of the earthquake and the resulting behaviour of the subsoil • The magnitude of the forces endured by the building in any given earthquake-induced ground motions • Amplitude, frequency, and duration of the ground motion As a result, extensive earthquake analyses can be done for a given site, depending on the importance and structural integrity of the structure. By assessing the hazards caused by the earthquake, it is possible to mitigate the effects of strong earthquakes, reducing the loss of life, injuries, and damages.   4.1  Seed’s Simplified Analysis Using Seed’s Simplified Equation developed by Seed and Idriss (1971), the simplified procedure can be used to estimate the cyclic shear stresses due to the earthquake for level sites. For a given depth in each soil layer, typically at the midpoint, a cyclic stress ratio can be calculated for a given magnitude earthquake, depending on the vertical effective stress at that given depth. Consequently, by plotting the values of tip resistance against cyclic stress ratio, the likelihood of  22liquefaction can then be identified. The following formula presents the calculation of the cyclic stress ratio (CSR):    where  amax = the peak ground surface acceleration for the design earthquake g = gravity acceleration ov = total vertical stress o’v = effective vertical stress rd = stress reduction factor  (rd = 1.0 – 0.00765z  for  z ≤ 9.15m rd = 1.174 – 0.0267z  for  9.15m ≤ z ≤ 23m) where z is the depth below ground surface, in metres  Using the chart below, comparing the corrected tip resistance measured from the CPT at each given depth with the calculated CSR, depending on where the point lies, it is then known that whether or not the soil layer will liquefy upon a Magnitude 7.5 earthquake:   Figure 4: Seed’s simplified liquefaction assessment with Corrected CPT Tip Resistance vs. CSR (obtained from Canadian Foundation Engineering Manual, p. 107)  23The table below illustrates the results of the liquefaction analysis from the CPT results:  Depth (m) From To Layer Thickness (m)(CSR)eqk qc1N (kg/cm²) Potential  for Liquefaction 0.00 1.00 1.00 0.162 6.61 Yes 1.00 3.75 2.75 0.109 8.47 Yes 3.75 5.25 1.50 0.180 74.23 Yes 5.25 7.00 1.75 0.193 130.65 No 7.00 13.00 6.00 0.203 48.15 Yes 13.00 14.75 1.75 0.189 24.93 Yes 14.75 19.00 4.25 0.174 56.30 Yes 19.00 23.00 4.00 0.151 72.43 Yes 23.00 23.75 0.75 0.136 54.64 Yes 23.75 27.75 4.00 0.121 137.82 No 27.75 29.25 1.50 0.103 40.97 Yes 29.25 30.25 1.00 0.095 77.89 No 30.25 32.25 2.00 0.085 33.73 Yes  Table 4: Liquefaction analysis of CPT#1 (Tip Resistance vs CSR)  As per the analysis, most soil layers are prone to liquefy upon an earthquake with a magnitude of 7.5, except at depths 6.13m, 25.75m, and 29.75m. At these depths, the soil layers are all sand, possessing high tip resistance values thus indicating high relative densities.  Depth (m) From To Layer Thickness (m)(CSR)eqk qc1N (kg/cm²) Potential for Liquefaction 0.00 2.00 2.00 0.129 24.36 Yes 2.00 3.75 1.75 0.153 193.43 No 3.75 8.75 5.00 0.157 415.33 No 8.75 13.00 4.25 0.200 35.64 Yes 13.00 13.50 0.50 0.192 42.64 Yes 13.50 14.75 1.25 0.188 33.77 Yes 14.75 16.00 1.25 0.182 68.82 Yes 16.00 22.00 6.00 0.163 31.80 Yes 22.00 24.75 2.75 0.137 44.99 Yes 24.75 25.50 0.75 0.126 66.03 Yes 25.50 28.25 2.75 0.115 48.67 Yes  2428.25 30.25 2.00 0.099 49.43 Yes  Table 5: Liquefaction analysis of CPT#2 (Tip Resistance vs CSR) The soil layers from CPT#2 are generally loose and potentially liquefiable, with the exception of 2 layers: at depths 2.88m and 6.25m, which are the wet sensitive fine-grained and the sand to silty sand layers respectively. These two layers contain very high tip resistance and N60 values, suggesting the high relative density nature of the soil. Otherwise, the results exhibit the other soil layers have low tip resistance values, therefore suggests the low relative density nature of the soil.  In addition to a comparison between corrected CPT tip resistance and the CSR to identify the risk of liquefaction, a comparison using the corrected SPT data between (N1)60 values and CSR can also be done to confirm the risk of liquefaction for a M = 7.5 earthquake. This chart is similar to the one shown in Figure 4:  Figure 5: Seed's simplified liquefaction assessment using (N1)60 values vs CSR  25Below are the results done for the liquefaction assessment using the SPT resistance method, for CPT#1: Depth Layer (N1)60 (CSR)eqk Potential for(m)       Liquefaction1 Sensitive Fine Grained 4.04  0.13  YES 2 Clay 9.87  0.13  YES 3 Clay 14.87  0.16  YES 4 Clay + Sand to Silty Sand 28.18  0.17  NO 5 Sand to Silty Sand 14.58  0.19  YES 6 Sand to Silty Sand + Sand 25.44  0.19  NO 7 Sand 18.48  0.20  YES 8 Sand to Silty Sand 21.89  0.20  NO 9 Sand to Silty Sand 10.71  0.21  YES 10 Sand to Silty Sand 10.76  0.22  YES 11 Sand to Silty Sand 12.24  0.21  YES 12 Sand to Silty Sand 12.61  0.21  YES 13 Sand to Silty Sand 10.75  0.21  YES 14 Silty Sand to Sandy Silt 8.85  0.20  YES 15 Silty Sand to Sandy Silt + Sand to Silty Sand 10.24  0.20  YES 16 Sand to Silty Sand 14.24  0.19  YES 17 Sand to Silty Sand 15.01  0.19  YES 18 Sand to Silty Sand 13.45  0.18  NO 19 Sand to Silty Sand 14.77  0.18  NO 20 Sand 13.68  0.17  YES 21 Sand 15.52  0.16  YES 22 Sand 15.99  0.16  YES 23 Sand 15.28  0.15  NO 24 Sand to Silty Sand + Sand 14.17  0.15  NO 25 Sand 16.53  0.14  NO 26 Sand 18.30  0.13  NO 27 Sand 7.64  0.13  YES 28 Sand + Sand to Silty Sand 14.33  0.12  NO 29 Sand to Silty Sand 12.63  0.11  NO 30 Sand to Silty Sand + Sand 12.48  0.11  NO 31 Sand + Sand to Silty Sand 12.27  0.10  NO 32 Sand to Silty Sand 9.65  0.09  YES Table 6: Liquefaction of CPT #1 ((N1)60 values vs CSR)  26Comparing the two methods of analyzing the liquefaction potential, it can be observed that very similar results are found. Most of the layers that contain silt are the layers that are most prone to liquefaction, and as well the lower depths are also found to have lower to none liquefaction potential. The following table shows the results from CPT#2: Depth Layer (N1)60 (CSR)eqk Potential for(m)       Liquefaction1.0  Sensitive Fine Grained (Dry) 1.86  0.13  YES 2.0  Sensitive Fine Grained (Dry) 8.93  0.13  NO 3.0  Sensitive Fine Grained (Wet) 26.46  0.16  NO 4.0  Sensitive F.G. (Wet) + Sand to Silty Sand 21.01  0.17  NO 5.0  Sand to Silty Sand 15.96  0.18  NO 6.0  Sand to Silty Sand 16.16  0.19  NO 7.0  Sand to Silty Sand 64.10  0.20  NO 8.0  Sand to Silty Sand 27.54  0.20  NO 9.0  Sand to Silty Sand + Silty Sand to Sandy Silt 11.39  0.20  YES 10.0  Silty Sand to Sandy Silt 7.02  0.21  YES 11.0  Silty Sand to Sandy Silt 10.71  0.21  YES 12.0  Silty Sand to Sandy Silt 7.18  0.21  YES 13.0  Silty Sand to Sandy Silt 10.08  0.21  YES 14.0  Sand to Silty Sand + Silty Sand to Sandy Silt 11.70  0.20  YES 15.0  Silty Sand to Sandy Silt + Sand to Silty Sand 11.90  0.20  YES 16.0  Sand to Silty Sand 13.75  0.19  YES 17.0  Silty Sand to Sandy Silt 9.41  0.19  YES 18.0  Silty Sand to Sandy Silt 10.06  0.18  YES 19.0  Silty Sand to Sandy Silt 10.30  0.18  YES 20.0  Silty Sand to Sandy Silt 10.55  0.17  YES 21.0  Silty Sand to Sandy Silt 10.24  0.17  YES 22.0  Silty Sand to Sandy Silt 11.86  0.16  YES 23.0  Sand to Silty Sand 14.21  0.15  NO 24.0  Sand to Silty Sand 11.35  0.15  NO 25.0  Sand to Silty Sand + Sand 12.58  0.14  YES 26.0  Sand + Sand to Silty Sand 13.77  0.13  YES 27.0  Sand to Silty Sand 12.74  0.13  YES 28.0  Sand to Silty Sand 11.16  0.12  NO 29.0  Sand to Silty Sand + Sand 17.34  0.11  NO  2730.0  Sand 10.35  0.11  YES Table 7: Liquefaction of CPT #2 ((N1)60 values vs CSR)  Once again, the results from the two different analysis presented similar results. Therefore, using both analysis methods, we can clearly identify the trouble zones in the subsurface, and these layers of concern will be taken into consideration when designing the foundation of the building.   285.0  FOUNDATION DESIGN OPTIONS There are two major types of foundations that can be used for support of the building: a shallow foundation, which would consist of a shallow footing in a variety of shapes, or a deep foundation, such as using timber and steel-pipe piles. Through the analyses of the soil properties of the site, these different foundations types can be considered. As well, ground improvement techniques will be investigated, as the site situates on relatively loose soils.   5.1  Site Preparation Due to the settlement induced by the structure in the shallow layers such as the clay layer in CPT#1 and the sensitive fine-grained layer in CPT#2, ground improvement techniques would need to be used to help limit the settlements. A few techniques will be investigated for this site, such as preloading, dynamic compaction, and vibro-compaction.  5.1.1 Preloading  Prior to the design of the foundation, settlement issues have to be considered, whether it was short-term or long-term. Therefore, to deal with these issues, preloading is one of the key options. From the two CPT boreholes, the ultimate settlement values are presented as follows:  Borehole Ultimate Settlement (mm) CPT#1 476.98 CPT#2 539.97 Table 8: Ultimate settlements from the two CPT boreholes. The layers that are of main concern due to settlement would be the clay layer in CPT#1 and the sensitive fine-grained layer in CPT#2. The two  29CPT boreholes vary significantly in properties, therefore the two retaining wall design structures are very different. Using block dimensions of 59” x 29.5” x 29.5”, and a surcharge consisting of sand with a unit weight of 18kN/m³. The preload time will be when the soil is 90% consolidated by the sand surcharge. The results are presented below: Borehole Time (months) Surcharge (kPa) Height of Wall (m) Number of Blocks CPT#1 13.2 60.17 3.34 5 CPT#2 37.2 31.97 1.78 3 Table 9: Number of retaining wall blocks required, without vertical drains  As presented above, the north side of the site would take much longer to reach 90% consolidation compared to the south side. However, this issue can be resolved if vertical drains are inserted in the impermeable layer. For instance, if drains are inserted one tenth of the thickness of the soil layer that is drained on both sides, settlements can be accelerated by up to 25 times. Thus, vertical drains is an option that can be installed in the horizontal direction during preloading, but in this scenario, due to inadequate data, the effects of installing vertical drains will not be investigated.  Consequently, for the south end of the site, since the soil will settle more and take longer to consolidate, 5 blocks will be used throughout, and along the north end of the building, 3 blocks will be used, shown in Figure 8 below. The rest of the calculations for the retaining wall design is presented in Appendix G.    30 Figure 6: Illustration of the number of retaining wall blocks required for the site.  Settlement gauges and piezometers will be installed at various places on the preload. This will monitor the settlement during preloading, removal of the preload, and the construction period of the structure.  However, this method has its advantages and disadvantages, as presented in the table below: Advantages Disadvantages - low cost - not possible if there are time constraints - preload material can be re-used as backfill material - disposal of fill material may be difficult if need to be transported out - quiet technique (no vibration/noise), good for the environment - post-construction settlement is relatively small - if site needs to be expanded after construction is complete, initial decision might not have been the most feasible option Table 10: Advantages and disadvantages of preloading.  3 block retaining wall 5 block retaining wall  315.1.2  Dynamic Compaction By performing dynamic compaction on the soil, using a free-falling heavy weight over the ground surface, the soil can be compacted for the use of shallow foundations. In addition, this low-cost and effective method can be used to reduce the liquefaction potential of loose soils.  Using a weight of 150 kilonewtons (kN) and a drop height of 5 metres, the compaction can influence a depth of up to 4.33 metres below ground surface. These impacts will be implemented in phases, in which the early phase is a high-energy phase, which are designed to improve deeper layers, followed by the low-energy phase, which are designed to densify the surficial layers. Compaction can be done in intervals of 3 metres apart in a grid, throughout the whole area of the proposed structure site, except within 10.8 metres within other existing structures. (Calculations in Appendix G) However, there are limitations to this method of ground improvement. Ground deformation and vibration can occur during compaction, as during the impact ground can buckle or deform due to the impact of falling weight. As well, the impact of the tamper on the ground can send waves into the ground, which will affect the nearby structures and the people living and working in them. Therefore, ground vibration of the peak particle velocities have to be monitored during this process, making sure it does not exceed 50mm/s to affect nearby residential structures. Finally, the efficiency of the process of improving clays and  32fine-grained material remains to be unproven, since during compaction, it creates an increase in pore water pressure.    5.1.3  Vibro-Compaction Processes This compaction method involves using an approximately 300mm to 400mm in diameter and a 4 to 5 metres long vibrator, which can either be electrically driven or hydraulically driven vibrator with variable frequency. Then the vibrator will penetrate the soil under its own weight, with water or airjetting and the induced vibration to assist with the densification of the soil. By doing so, it can reduce the volume of the soil by up to 10%, thus, the level of the site might be altered by this process, therefore granular material can be placed around the vibrator.  The centre-to-centre spacings of this stone column method can be done in spacings of 1.5 metres, to achieve compaction of the soils of up to 90%. This method of soil improvement has four basic objectives: - to limit total settlements - to reduce differential settlements - to achieve higher bearing capacity - to increase shear strength (CFEM, p.250)   335.2  Shallow Foundations Shallow foundations can provide building loads to the earth at a shallow depth, usually consists of spread footings and mats/raft foundations. This category of foundation has a few advantages: low cost, simple to construct, uses mostly concrete (do not need to use many different types of materials), and easier labour (no need to do as many inspections as deep foundations). However, there are a few disadvantages as well, such as: settlement, foundation failures such as bearing capacity failures, punching failures, and slope failures in certain soil conditions.  5.2.1 Strip/Spread Footings Analysis of the strip and spread footings were done for the following shaped footings: square, circular, continuous, and rectangular. Two methods were used to investigate the bearing capacity and allowable column load of the footing: the Terzaghi method, which is based on a general shear failure, and the Vesic (also known as the Meyerhof method), which includes correction factors for eccentricity, load inclination, and foundation depth. Contrary to the simpler Terzaghi method, the Vesic method includes the influence of shear strength above the base of the foundation and provides a more accurate bearing values and it applies to a much broader range of loading and geometry conditions. (Coduto, p.183) However, for our analysis, to be conservative and using a Factor of Safety value of 3.0, the Vesic values will be considered in the design since it provides a lower bearing capacity values between the two methods, since  34this will induce a bigger settlement values for the settlement analysis methods.  For the analysis of the shallow foundation, the data from the CPT results will be used. From these two sets of data and results, the dimensions of the footings can be determined, along with the bearing capacity and allowable column load. From the structural engineer, it was given that the column loads (dead and live loads) were 200 kips (889.64kN) and the maximum floor live loads (dead and live loads) were 400 pounds per square feet (psf), or 19.152kPa. Therefore, our design considerations were taken in accordance to these numbers.  The following dimensions and depths of the footings will provide adequate bearing capacities and column loads for the building. Below are the results from CPT#1:   Square Circular Continuous RectangularWidth (m) 2 2.5 3 1 Length (m) --- --- --- 4 Depth (m) 1.5 1.5 2 1.5 qult (kPa) 1063 1056 1027 837 qa (kPa) 354 352 342 279 Allowable Column Load (kN) 1418 1728 1027 1115 Table 11: Footing dimensions and capacities from CPT#1     For CPT#2:   Square Circular Continuous RectangularWidth (m) 2 2.3 2.5 1 Length (m) --- --- --- 2.5 Depth (m) 1 1 2 2 qult (kPa) 978 988 1281 1422 qa (kPa) 326 329 427 474  35Allowable Column Load (kN) 1305 1368 1068 1185 Table 12: Footing dimensions and capacities from CPT#2  From the allowable column load values obtained from the bearing capacity calculations, resulting maximum settlement and bearing capacity values for the site conditions can then be attained. The Classical method (conventionally known as the Terzaghi method), will be used in this analysis. This theory assumes settlement is a one-dimensional process, in which all strains are vertical. (Coduto, p.218) The results from the Classical method are presented below: From CPT #1:   Square Circular Continuous RectangularWidth (m) 2 2.5 3 1 Length (m) --- --- --- 4 Depth (m) 1.5 1.5 2 1.5 Allowable Column Load (kN) 1418 1728 1027 1115 q (kPa) 390 388 390 314 delta (mm) 308.09 305.46 269.34 253.87 Table 13: Footing load and settlement values for CPT #1 (Classical Method)  From CPT #2:   Square Circular Continuous RectangularWidth (m) 2 2.3 2.5 1 Length (m) --- --- --- 2.5 Depth (m) 1 1 2 2 Allowable Column Load (kN) 1305 1368 1068 1185 q (kPa) 350 353 474 521 delta (mm) 396.76 398.83 332.73 292.13 Table 14: Footing load and settlement values for CPT #2 (Classical Method)  36Settlement values from these given foundations are noticeably large, greatly exceeding the values given from the guidelines for limiting settlement of framed buildings and load bearing walls, expressed in CFEM page 180. The angular distortion is a ratio between the settlement and the span between columns, and the tolerable limit of this value for a shallow foundation is between 1/150 to 1/250. That means for every 150 to 250 millimetre span between columns, 1 mm of settlement is allowed. The span between columns is determined by the structural engineers. Comparing the allowable maximum settlement values with the calculated settlement values from the four different shapes of foundations using two different computation methods, it is necessary to preload the site prior to any construction of the shallow foundations. The preloading requirements and methods are presented in Section 5.1.1.  5.2.2  Mat Foundation Since the footprint of the building is relatively large, a mat, also known as a raft, foundation can be considered. Columns and walls of the building would be supported by this common foundation, and this type of foundation is ideal for reducing or distributing building loads in order to reduce differential settlements for surrounding areas.  However, since the building footprint is rather large, different soil conditions will be present throughout the site. Therefore when obtaining the value of coefficient modulus of subgrade reaction of the subsurface soils, the value will only be an approximation. With disturbances of the  37soils in events such as excavation and placement of foundations, the coefficient modulus will be presented in a range of values for a specific soil type. They are as follows: Soil Type Kv1 (MPa/m) Granular Soil (Moist or Dry)  Loose 5 – 20 Compact Sand 20 – 60 Dense 60 – 160 Very Dense 160 – 300 Cohesive Soils  Soft < 5 Firm 5 – 10 Stiff 10 – 30 Very Stiff 30 – 80 Hard 80 – 200 Table 15: Typical subgrade reaction values (Table obtained from CFEM, p.129, Table 7.1)  Through classification of the top soil layers, it is determined that the granular soil has a Kv1, which is the modulus of subgrade reaction for one-foot square plate, of 20 MPa/m. In addition, since the maximum floor load is determined to be 400 psf (equivalent to 19.152 kPa), we are then able to obtain the value of settlement of footing under applied pressure, δ, through this formula: k = q / δ Using the k-value of 20 MPa/m, we have to convert this modulus to the suitable actual footing dimensions, b. Therefore, we use the following formulas to obtain a foundation width and a modulus for the actual footing dimension b for granular soils: kvb = kv1 [(b+1)/2b]²  38δ = Iqb(1-υ²) / E where I = an influence factor that is dependant on geometry of footing and thickness of compressive soil relative to footing (value is 1, according to CFEM p.161) b = foundation width υ = Poisson’s ratio (typically 0.3 for drained conditions for most soils) E = modulus of deformation   The following are the calculations to find the value of the foundation width, b, and the amount of settlement induced by the raft: δ 0.174557 m   174.5566 mm b 0.263078 m Table 16: Mats foundation results from CPT #1   δ 0.024137 m   24.13674 mm b 0.657696 m Table 17: Mats foundation results from CPT #2  Firstly, the bearing capacities of the foundation were found from the data of the two CPT boreholes, using a Factor of Safety value of 3. Data is presented as follows:       Terzaghi Vesic     Bearing Capacity Allowable Wall Load Bearing Capacity Allowable Wall Load   Width (B) (m) Depth (D) (m) qult (kPa) qa (kPa) P/b (kN/m) qult (kPa) qa (kPa) P/b (kN/m)CPT #1 0.263078 0.5 243 81 21 263 88 23CPT #2 0.657696 0.5 312 104 68 325 108 71Table 18: Calculated bearing capacity values for mats foundation  To be conservative, the comparatively lower values will be considered for the calculations settlement analysis.   39The Classical Method was used to calculate the settlement induced by the raft foundation from data of the two CPT boreholes.  Treating this foundation as a continuous strip footing, with a depth of 0.5m and using allowable wall load from Terzaghi, along with the foundation width obtained from the previous table, the maximum load values can be obtained, as displayed below:  q (maximum allowable load) (kPa) Total Settlement (mm) Classical Method   CPT #1 92 106.83 CPT #2 115.191 192.62 Table 19: Calculated settlement values for mats foundation  As mentioned previously, the maximum live and dead floor load values of the building provided by the structural engineer is said to be 400 psf (equivalent to 19.152 kPa). Therefore, using data from both settlement methods and CPT boreholes, a Factor of Safety value of approximately 5 and 6 would be achieved for CPT borehole 1 and 2 respectively. This greatly exceeds the Factor of Safety requirement of 3, typical in design factors. However, with such large settlement values from the given widths and depths of the foundation, preloading is necessary to reduce the amount of induced settlement from the structure.    405.3  Deep Foundations A deep foundation with piles can also be considered for this proposed building. Since the upper soil layers of this site are rather weak, pile foundations such as timber piles or steel piles are considered when designing for the foundation. These piles can provide a large load capacity because of strong mid- and lower-layered soils, and it is good for seismically active areas such as the location of this proposed site.   5.3.1  Timber Piles Timber piles are a good and economical choice for a foundation, since they are made from trunks of straight trees, and trees are abundant in the province of British Columbia. This material is typically tapered, and usually coming from Southern pine or Douglas fir in North America. After the bark and branches are removed from these trees, they can be driven 20 to 60 feet into the ground.  In addition, lengths of up to 80 feet for Southern pine and 125 feet for Douglas fir can be obtained. (Coduto, p.379-380)  Conversely, timber piles are very susceptible to damage during driving. Repeated blows by the driving hammer can induce damages which include splitting and brooming at the head and the toe. Even though preventative measures such as using a lighter weight hammer, inserting steel shoes on the toe of the pile, and predrilling prior to installation of the timber piles, it is not necessarily sufficient to prevent damage of the piles. In addition, and the most important point on this given proposed site,  41timber piles are not usually good for hard and dense soils, as present at the lower-layered soils at this site. (Coduto, p.381) Finally, timber piles typically have an axial design loads of 35-150 kips, but the structural engineers determined that the live and dead loads of the building is 200 kips, therefore this pile type is proven to be inadequate for this building. (Timber Pile Manual, p.14)   5.3.2  Steel Piles Steel piles are a popular choice for foundation, given its high bearing capacity capabilities and is ideal for dense and hard soil conditions. There are two types of steel piles: H-piles, and pipe piles. Pipe piles will be considered in the design since pipe piles provide larger lateral loads, needed for this site since the great risk of seismic activity is present.  There are two types of steel piles: open-end, and closed-end. In this scenario, only closed-end piles are considered since open-ended piles are primarily for offshore construction. In addition, closed-end pipe piles have higher load capacities than open-end ones.  For our analysis of a steel pile design, a free program called Louisiana Pile design by Cone Penetration Test developed by the Louisiana Transportation Research Center will be primarily used for the analysis of the piles. This program only allows for the pile material to be concrete, so in order to obtain the pile bearing capacities for steel, a 10% reduction from the concrete pile’s bearing capacity will be implemented.  42This is because steel has a less frictional sliding factor than concrete, therefore the ability for the soil to hold the friction pile up would decrease the bearing capacity by 10%.  In this program, there are three types of analysis for pile capacities based on the CPT data: the LCPC method, Schmertmann method, and de Ruiter and Beringen method. The LCPC method, which is the most popular method in industry standards, relies on the tip resistance averaged over zone above and below the tip of the cone to get an Equivalent Cone Resistance, qca. Thereafter, adding the side friction and tip resistance, which uses the equivalent cone resistance to calculate, an ultimate capacity for the pile can be obtained. The Schmertmann method divides each layer into approximately equal or representative values of qc. Then a pile group is represented as a raft, and that would be superimposed at each given depth. There, pile capacities can be calculated at each depth. (Pile Design and Construction Practice, p.185) Finally, the de Ruiter and Beringen method have different procedures for clays and sands. In clays, the friction and end bearing capacities rely on calculating the undrained shear strength, su, first from cone resistance, qc. In sands, the pile end bearing, qp, is calculated using the cone resistance in a zone 0.7 to 4 pile diameters below the pile tip. As well, pile capacities in overconsolidated sands may be partially reduced due to pile driving, therefore making it hard to obtain an exact value. (CPT in Geotechnical Practice, p.154)   43An analysis using three different pile sizes, 10-inch, 12-inch, and 16-inch, is used to calculate the bearing capacity of the piles using the aforementioned three analysis methods. Below are the results from CPT 1, with capacities expressed in tons (metric): Pile Diameter Depth (feet) LCPC Schmertmann de Ruiter and Beringen 20 21.9 11.9 13.5 30 39.5 17.5 21.8 40 53.5 22.4 29.9 50 67.3 28.6 39.4 60 85.1 35.0 46.3 70 107.3 42.6 54.3 80 129.7 53.1 63.5 90 150.0 65.1 74.8 10 100 171.6 78.9 85.0 20 26.3 14.9 16.2 30 47.4 21.6 26.1 40 64.2 27.5 35.9 50 80.7 34.9 47.2 60 102.1 42.6 55.6 70 128.8 51.8 65.2 80 155.7 64.3 76.2 90 180.0 78.7 89.8 12 100 205.9 95.3 102.1 20 35.1 21.4 21.5 30 63.1 31.0 34.8 40 85.6 38.7 47.8 50 107.7 48.7 63.0 60 136.1 59.0 74.1 70 171.8 71.2 86.9 80 207.6 87.9 101.7 90 240.0 107.1 119.7 16 100 274.5 129.1 136.1 Table 20: Pile capacity values from CPT#1 for steel piles  Below are the values calculated from CPT 2: Pile Diameter Depth (feet) LCPC Schmertmann de Ruiter and Beringen 20 29.2 50.5 35.3 30 49.6 78.4 60.8 40 65.9 107.1 77.9 50 86.2 134.9 103.9 60 106.7 163.7 128.0 10 70 126.7 191.5 153.5  4480 148.8 220.3 182.2 90 170.7 248.1 210.0 20 35.0 60.7 42.4 30 59.5 94.0 72.9 40 79.1 128.5 93.5 50 103.5 161.9 124.7 60 128.0 196.4 153.6 70 152.1 229.8 184.2 80 178.6 264.3 218.6 12 90 204.8 297.7 252.0 20 46.7 80.9 56.5 30 79.3 125.4 97.2 40 105.4 171.4 124.6 50 138.0 215.9 166.2 60 170.7 261.9 204.8 70 202.8 306.4 245.6 80 238.1 352.4 291.5 16 90 273.1 396.9 336.0 Table 21: Pile capacity values from CPT#2 for steel piles  Since the loads (dead and alive) value provided by the structural engineers is 200 kips - equivalent to 90.718474 tons (metric), the values represented in red indicates the pile capacity values that were higher than the required loads. Since the LCPC method provides the smallest bearing capacities of the three methods, only values obtained from the LCPC method will be considered.   5.3.3  Concrete Piles  Another choice for a pile foundation could be using piles made from concrete. They usually come in two different forms: an in-situ form for bored piles and cylinders, or in precast form for driven piles in either reinforced or prestressed concrete. (Young, p.194) The in-situ form of concrete piles requires pouring concrete into a preformed hole or driven  45tube into the ground, with temporary or permanent steel lining tubes which will provide support for the unhardened concrete mix. It will then be left for 28 days before any testing will be done on the piles. (Young, p.196) Reinforced concrete piles are required to either have a steel or plastic reinforcement surrounding the pile. Newer technologies of concrete piles include injecting a polypropylene fibre into the concrete mix to provide the concrete with better strength and lower costs. However, this alternative will provide lower Young’s modulus. (Young, p.198) Prestressed concrete piles require a certain amount of stress induced on the pile before being placed on the ground. However, these piles will need to be handled very carefully before placing into the ground. As well, during driving of these piles, adequate cushioning material must be needed between the driving head and the concrete pile; otherwise the pile is very prone to breakage. This cushioning material can be made of plywood, and it will situate between the hammer head of the pile driving rig, which will also contain a hammer cushion, and the concrete pile itself.  Using concrete for piles allows the easy adjustments of the concrete mixing material for different environments and usages of these piles. However, even though concrete piles are a cheaper and provide a more adaptive to the environment alternative to steel piles, one of its great disadvantages is its low shear strength. That means when large lateral loads are acted on the pile, it is very prone to failure. (Young, p.194)   46Below are values obtained from both CPT #1 and #2. As described previously, the values will be 10% greater than the bearing capacities from steel: Pile Diameter Depth (feet) LCPC Schmertmann de Ruiter and Beringen 20 24.4 13.3 15.0 30 43.9 19.5 24.2 40 59.5 24.9 33.2 50 74.8 31.8 43.7 60 94.5 38.9 51.5 70 119.3 47.4 60.4 80 144.2 59.0 70.6 90 166.6 72.3 83.1 10 100 190.6 87.6 94.5 20 29.3 16.5 18.0 30 52.6 24.1 29.0 40 71.4 30.5 39.8 50 89.7 38.8 52.5 60 113.4 47.4 61.8 70 143.1 57.5 72.4 80 173.0 71.4 84.7 90 200.0 87.5 99.7 12 100 228.7 105.8 113.4 20 39.0 23.8 23.9 30 70.2 34.4 38.7 40 95.1 43.0 53.1 50 119.6 54.1 70.0 60 151.2 65.5 82.4 70 190.8 79.1 96.6 80 230.7 97.6 112.9 90 266.6 119.0 133.0 16 100 305.0 143.5 151.2 Table 22: Pile capacity values from CPT#1 for concrete piles  47 Pile Diameter Depth (feet) LCPC Schmertmann de Ruiter and Beringen 20 24.4 13.3 15.0 30 43.9 19.5 24.2 40 59.5 24.9 33.2 50 74.8 31.8 43.7 60 94.5 38.9 51.5 70 119.3 47.4 60.4 80 144.2 59.0 70.6 90 166.6 72.3 83.1 10 100 190.6 87.6 94.5 20 29.3 16.5 18.0 30 52.6 24.1 29.0 40 71.4 30.5 39.8 50 89.7 38.8 52.5 60 113.4 47.4 61.8 70 143.1 57.5 72.4 80 173.0 71.4 84.7 90 200.0 87.5 99.7 12 100 228.7 105.8 113.4 20 39.0 23.8 23.9 30 70.2 34.4 38.7 40 95.1 43.0 53.1 50 119.6 54.1 70.0 60 151.2 65.5 82.4 70 190.8 79.1 96.6 80 230.7 97.6 112.9 90 266.6 119.0 133.0 16 100 305.0 143.5 151.2        Table 23: Pile capacity values from CPT#1 for concrete piles  Outlined in red are the values that exceed the building load of 200 kips (90.718474 metric tons). We will also consider the LCPC method as the main method of analysis for concrete piles. In addition, the same pile diameters of 10, 12, and 16-inches will be considered.    486.0  THE CONSTRUCTION PROCESS 6.1  Vibration Monitoring  During the pile driving process, and depending on the diameter of the pile being driven, ground vibration was felt throughout and outside the construction site. Therefore, around the vicinity of the construction site, vibration monitoring tests using a seismograph were performed to make sure the vibration caused by the pile driving process would not affect the structural integrity of the nearby residential and commercial buildings. These tests were done from a few feet away from the driven pile to as far as 160 feet away. According to the British Columbia Building Code 2006, the peak particle velocity (PPV) which is the maximum allowable vibration for residential complexes is 50 millimetres per second (mm/s) and for commercial complexes is 100mm/s. The table and figure below demonstrates the amount of PPV induced from the pile driving relative to the distance away from the driven pile: Distance (ft) Average PPV (mm/s) Max PPV (mm/s) Min PPV (mm/s) 20 8.345 10.4 6.1 40 8.5575 11.7 7.11 80 5.232 6.86 4.03 120 2.2 3.05 2.03 160 1.46 1.9 1.02         Table 24: Vibration monitoring data  49Distance vs PPV024681012140 20 40 60 80 100 120 140 160 180 200Distance (ft)PPV (mm/s)Average PPV Maximum PPV Minimum PPV Figure 7: PPV relative to the distance away from the source  These data were collected at random times when different diameter piles were driven, therefore only upper and lower-bound values were plotted, as well as the average of the values obtained. The closest residential and commercial complexes were at least 100 feet away from the construction site, therefore from the plot above it proves that none of the surrounding structures were affected by the pile driving process.    In addition, vibration tests were done within 20 feet of the driven pile as well. This was done to ensure that the adjacent commercial complexes will not experience structural damage during pile driving. Values obtained from the monitoring are shown below:  Driving Energy (ft-lb) Hammer Drop (ft) Distance (ft) PPV (mm/s) 3 115 4 72 10 28 50,000 8 15 20 Table 25: Vibration monitoring close to the source   50Since the typical driving energy of the hammer was 50,000 ft-lb, then using the largest hammer drop of 8 feet, it was found that only when the seismograph was placed 3 feet away from the pile that the PPV exceeded the commercial building limit of 100 mm/s. However, the closest commercial building to any pile placed was at least 10 feet away, therefore the vibration induced by the pile driving did not affect the structural integrity of the adjacent commercial buildings.  Distance vs. Peak Particle Velocity0204060801001200 50 100 150 200 250Distance (ft)PPV - Peak Particle Velocity (mm/s)Average PPV Maximum PPVMinimum PPVDamage ThresholdDamage Threshold for Residential Buildings (at 50 mm/s)Driving Energy at 50,000 ft-lbDamage  Thre s hold for Comme rcial Buildings  (at 100 mm/s ) Figure 8: Distance vs Peak Particle Velocity values relative to damage threshold values  516.2  PDA Testing and Checking Integrity of Piles To ensure the driven pile is performing as expected, pile driving analyzer tests were performed on the pile while it was being driven. The Case Pile Wave Analysis Program (CAPWAP) is a combination of the Case method, which uses wave trace data to determine the static pile capacity, and the wave equation analysis, which uses a much more precise numerical model but does a weak estimate of the actual energy delivered by the hammer. Producing the values of ultimate resistance in the soil “springs”, as well as the quake and Case method damping factor, CAPWAP measures the total capacity, along with the shaft and toe resistances of the pile. In addition, with the use of the combination of an accelerometer and strain gage during a PDA test, the axial forces of the pile as well as the particle velocity of the waves travelling through the pile and the pile displacement during the hammer blow can be calculated. By obtaining these data, we can then produce a plot of time versus particle velocity to check the integrity of the pile, ensuring the pile did not break or fracture during the pile driving process, or plotting time versus force to find out whether the pile undergoes compressive or tensile forces while the pile is being driven. This analysis allows the simulation of a static load test.    PDA tests for the Shoppers Drug Mart site were done on June 19th, 2007 and July 3rd, 2007 by AATech Scientific Inc. based in Ottawa, Ontario. This analysis was done on a total of 10 randomly selected test piles, varying in pile diameter and location. The values were obtained using CAPWAP® Version 2000-1. Using a 6300 pound hammer, with height of hammer drop varying from 5  52to 8 feet, values were obtained from these tests. A summary of the values obtained from the tests are displayed below:  Date Pile Number Pile Diameter (in) Gridline Total Capacity (kips) Allowable Capacity (kips) Factor of Safety 19-Jun-07 3067 10 1A 192.9 100 1.929 19-Jun-07 3091 16 1A 165.1 100 1.651 19-Jun-07 3091 16 1A 178.4 100 1.784 19-Jun-07 3100 16 1 144.9 100 1.449 19-Jun-07 3113 16 2 180.1 100 1.801 19-Jun-07 3207 12 4 127.2 100 1.272 19-Jun-07 3230 12 4 139.9 100 1.399 03-Jul-07 3064 10 1 185 100 1.85 03-Jul-07 3079 12 1A 187.2 100 1.872 03-Jul-07 3097 16 1 205 100 2.05 03-Jul-07 3100 16 1 163.4 100 1.634 03-Jul-07 3113 16 2 204.6 100 2.046 03-Jul-07 3120 16 2 195 100 1.95 Table 26: Final tested pile capacity values  As per the table above, as time progresses, we can see the pile increase in total capacity. When the first PDA tests were done for the piles on June 19th, 2007, the piles were only placed in approximately 3 days before the testing, therefore the piles did not have enough time to settle into the soil stratums. However, since the surrounding soil layers around the driven piles are normally consolidated, it disturbs and remolds the surrounding soil, therefore generating excess pore water pressures. But with time, through the settling of the piles and the dissipation of pore water pressure of the soil around the pile as a result of soil consolidation, the pile load carrying capacity increases over time. Therefore, several more PDA tests were done on the pile approximately 2 weeks after installation of the piles, some of which were re-tested to observe this “pile aging” affect of the piles.    537.0  RESULTS AND CONCLUSION  Preloading for the site would not be a practical choice for ground improvement for the site, because of two main reasons: firstly, it will require too much time for the preload (approximately 13.2 and 37.2 months for north and south ends of site). Secondly, and the most important reason, is after seismic design considerations, it was found that the below soil stratums contain many layers that will liquefy upon an earthquake, it is then necessary to perform soil densification measures. Therefore, the best solution for ground improvement would be to use vibro-compaction. To improve the quality of the subsurface, it is then necessary to have centre-to-centre spacings of 1.5 metres.  For the foundation of the building, it is determined that shallow foundation by itself will not be adequate, since the subsurface soils are very prone to liquefaction upon an earthquake. In addition, other than a mats foundation, settlements will be a large issue for the shallow foundation as well, mainly because of the upper weak soils. Therefore, a deep foundation is our only option for the foundation, to ensure that the foundation reaches a layer of soil that will not liquefy.  After considering the three different pile foundation types, it is found that steel piles are best suited for the given environment of the proposed site. Even though timber piles are the most cost effective choice, since the soil layers below the depth of 15 to 20 feet are relatively dense and hard soils, timber piles are prone to split or break apart while pile driving in such scenarios. In addition, the given dead and live loads exceed the typical bearing capacity of this type of pile.  54Concrete piles are a cheap alternative as well, but since as stated previously, one of its greatest disadvantages is that it has a low shear strength. Since this proposed site is situated in a seismically active area, it would be very risky to place concrete piles as foundation, as in such events the piles are prone to breaking apart, causing the foundation of the building to fail. Therefore, a pile foundation made of steel would be the best choice out of all, since steel has a higher shear strength. In a case of an earthquake, the piles are less prone to breaking apart. In addition, since the soil stratum of the site contains relatively harder material, it is more suitable to use a higher strength material such as steel, allowing the piles to not break while driving through such hard soils.  Using the proper diameter of the pile and determining the amount of piles needed in a pile group at certain locations is up to the discretion of the structural engineer. A proper length of the pile to use should be 60 feet, since at such depth the steel piles would not need to be spliced and would provide an adequate amount of bearing capacity. Splicing of the pile will cause the piles to fail at the spliced joints in the event of an earthquake, therefore splicing would not be an option for this foundation.   55LIST OF REFERENCES  Armstrong, John E. Vancouver Geology. Canada: Geological Association of  Canada, 1990, p.42-3.   Canada. Geological Survey of Canada. Catalogue of Canadian Volcanoes –  Garibaldi Volcanic Belt: Garibaldi Lake Volcanic Field. 13 Feb. 2008  < http://gsc.nrcan.gc.ca/volcanoes/cat/feature_garibaldi_e.php>  Canadian Geotechnical Society. Canadian Foundation Engineering Manual 4th  Edition. Richmond, B.C.: The Canadian Geotechnical Society, 2006, p.107, 129, 161, 250.  Coduto, Donald P. Second Edition Foundation Design – Principles and Practices.  Upper Saddle River, New Jersey: Prentice Hall, 2001, p.379-381.  Coduto, Donald P. Geotechnical Engineering – Principles and Practices. Upper  Saddle River, New Jersey: Prentice Hall, 1999.  Collin, James G. Timber Pile Design and Construction Manual. USA: American  Wood Preservers Institute, 2002, p.14  Lunne, T., J.J.M. Powell, and P.K. Robertson. Cone Penetration Test  Geotechnical Practice. London, England: Taylor & Francis, 1997, p.154.   Mathews, Bill and Jim Monger. Roadside Geology of Southern British Columbia.  Missoula, Montana: Mountain Press Publishing Company, 2005, p.162-3.   Mathews, William H. Garibaldi Geology: A popular guide to the geology of the  Garibaldi Lake area. Vancouver, British Columbia: Geological Association of Canada, 1975.   Monger, J.W.H. Geological Survey of Canada Bulletin 481: Geology and  Geological Hazards of the Vancouver Region, Southwestern British Columbia. Ottawa, Ontario: Canada Communications Group, 1994, p.11-3, 221-232, 239, 267-9.   Squamish (B.C.) District Council. Squamish Official Community Plan. N.p. The  District of Squamish B.C., 1989, p.12-13, 24-26, 29-30, 34-36, 42, 85.   Stathers, Jack Kenneth. A Geographical Investigation of Development Potential  in the Squamish Valley Region, British Columbia. Vancouver: The University of British Columbia, 1958, p.19-21, 60-61.  Tomlinson, Micha J. Pile Design and Construction Practice. London, England:   56E&FN Spon, 1994, p.185.   Ulrich, Edward J. Design and Performance of Mat Foundations. Detroit, Michigan:  American Concrete Institution, 1995.  Young, F.E. Piles and Foundation. London, England: Thomas Telford Ltd., 1981,  p.194, 196, 198.                     APPENDIX A: SITE PLAN                     APPENDIX B: REGIONAL GEOLOGY MAPS                     APPENDIX C: AERIAL PHOTOGRAPHS                     APPENDIX D: AUGER DRILL DATA                     APPENDIX E: CPT DATA                     APPENDIX F: CROSS SECTION                     APPENDIX G: RESULTS AND CALCULATIONS  Project Number: Title: Date:CPT No: Parish: Station:      Offset: Elevation (ft): 01020304050607080901000 50 100 150 200Depth (ft) Tip Resistance (ton/ft01020304050607080901000 1 2 3 45 Sleeve Resistance (ton/ft01020304050607080901000 2 4 6 8 10 Sleeve/Tip Ratio (%) % Sand % Silt% Clay0 20 40 60 80 1000 20 40 60 80 100 Probability of Soil TypesProject Number: Title: Squamish Shoppers Drug Mart Date:CPT No: 2 Parish: Station:      Offset: Elevation (ft): 01020304050607080900 100 200 300 400Depth (ft) Tip Resistance (ton/ft01020304050607080900 1 2 3 4 Sleeve Resistance (ton/ft01020304050607080900 1 Sleeve/Tip Ratio (%) % Sand % Silt% Clay0 20 40 60 80 1000 20 40 60 80 100 Probability of Soil TypesCPT #1Unit Weight Thickness σz0σ'z0From To Layer (kN/m³) (m) (kPa) (kPa)0.00 1.00 Sensitive Fine Grained 17.5 1.00 8.75 7.001.00 3.75 Clay 18.0 2.75 42.25 49.683.75 5.25 Sand to Silty Sand 19.0 1.50 81.25 56.755.25 7.00 Sand 19.5 1.75 112.56 72.147.00 13.00 Sand to Silty Sand 19.0 6.00 186.63 108.2313.00 14.75 Silty Sand to Sandy Silt 18.5 1.75 259.81 143.4414.75 19.00 Sand to Silty Sand 19.0 4.25 316.38 170.6019.00 23.00 Sand 19.5 4.00 395.75 209.5523.00 23.75 Sand to Silty Sand 19.0 0.75 441.88 232.4023.75 27.75 Sand 19.5 4.00 488.00 255.2527.75 29.25 Sand to Silty Sand 19.0 1.50 541.25 281.5529.25 30.25 Sand 19.5 1.00 565.25 293.3030.25 32.25 Sand to Silty Sand 19.0 2.00 594.00 307.35** GWT at 1.83-2.44m Chosen = 2.0mLiquefaction AssessmentLayer Thickness (m)Depth at Middle (m)σz0 (kPa) σ'z0 (kPa) rd(CSR)eqkqc (kg/cm²)qc1N (kg/cm²)Potential for Liquefaction1.00 0.50 8.75 7.00 0.996 0.162 1.78 6.61 Yes2.75 2.38 42.25 49.68 0.982 0.109 6.09 8.47 Yes1.50 4.50 81.25 56.75 0.966 0.180 57.07 74.23 Yes1.75 6.13 112.56 72.14 0.953 0.193 113.26 130.65 No6.00 10.00 186.63 108.23 0.907 0.203 51.13 48.15 Yes1.75 13.88 259.81 143.44 0.804 0.189 30.47 24.93 Yes4.25 16.88 316.38 170.60 0.723 0.174 75.06 56.30 Yes4.00 21.00 395.75 209.55 0.613 0.151 107.02 72.43 Yes0.75 23.38 441.88 232.40 0.550 0.136 85.02 54.64 Yes4.00 25.75 488.00 255.25 0.486 0.121 224.73 137.82 No1.50 28.50 541.25 281.55 0.413 0.103 70.16 40.97 Yes1.00 29.75 565.25 293.30 0.380 0.095 136.14 77.89 No2.00 31.25 594.00 307.35 0.340 0.085 60.35 33.73 YesFriction AngleDepth qc (MPa) qc (kg/cm²) σ'z0 (kPa)Friction Angle (φ)0.50 0.175 1.78 7.00 302.38 0.598 6.09 49.68 304.50 5.597 57.07 56.75 426.13 11.107 113.26 72.14 4410.00 5.014 51.13 108.23 3713.88 2.988 30.47 143.44 3216.88 7.361 75.06 170.60 3621.00 10.495 107.02 209.55 3723.38 8.338 85.02 232.40 3525.75 22.039 224.73 255.25 4028.50 6.880 70.16 281.55 3229.75 13.351 136.14 293.30 3631.25 5.918 60.35 307.35 30SandSand to Silty SandSoil LayerDepth (m)Sensitive Fine GrainedClaySand to Silty SandSilty Sand to Sandy SiltSand to Silty SandSandSand to Silty SandSand to Silty SandSandSand to Silty SandSandCPT #2Unit Weight Thickness σz0σ'z0From To Layer (kN/m³) (m) (kPa) (kPa)0.00 2.00 Sensitive Fine Grained (Dry) 17.5 2.00 17.50 17.502.00 3.75 Sensitive Fine Grained (Wet) 18.5 1.75 51.19 42.603.75 8.75 Sand to Silty Sand 19.5 5.00 116.13 91.638.75 13.00 Silty Sand to Sandy Silt 18.5 4.25 204.19 117.2113.00 13.50 Sand to Silty Sand 19.0 0.50 248.25 138.0013.50 14.75 Silty Sand to Sandy Silt 18.5 1.25 264.56 145.7414.75 16.00 Sand to Silty Sand 19.0 1.25 288.00 156.9316.00 22.00 Silty Sand to Sandy Silt 18.5 6.00 355.38 188.7822.00 24.75 Sand to Silty Sand 19.0 2.75 437.00 227.5324.75 25.50 Sand 19.5 0.75 470.44 243.8125.50 28.25 Sand to Silty Sand 19.0 2.75 503.88 260.1028.25 30.25 Sand 19.0 2.00 549.00 281.95** GWT at 1.83-2.44m Chosen = 2.0mLiquefaction AssessmentLayer Thickness (m)Depth at Middle (m)σz0 (kPa) σ'z0 (kPa) rd(CSR)eqkqc(kg/cm²)qc1N (kg/cm²)Potential for Liquefaction2.00 1.00 17.50 17.50 0.992 0.129 10.40 24.36 Yes1.75 2.88 51.19 42.60 0.978 0.153 128.86 193.43 No5.00 6.25 116.13 91.63 0.952 0.157 405.76 415.33 No4.25 10.88 204.19 117.21 0.884 0.200 39.39 35.64 Yes0.50 13.25 248.25 138.00 0.820 0.192 51.13 42.64 Yes1.25 14.13 264.56 145.74 0.797 0.188 41.61 33.77 Yes1.25 15.38 288.00 156.93 0.763 0.182 87.99 68.82 Yes6.00 19.00 355.38 188.78 0.667 0.163 44.59 31.80 Yes2.75 23.38 437.00 227.53 0.550 0.137 69.26 44.99 Yes0.75 25.13 470.44 243.81 0.503 0.126 105.23 66.03 Yes2.75 26.88 503.88 260.10 0.456 0.115 80.11 48.67 Yes2.00 29.25 549.00 281.95 0.393 0.099 84.72 49.43 Yes** Assume Peak Ground Acceleration (g) = 0.2gFriction AngleDepth qc (MPa) qc (kg/cm²) σ'z0 (kPa)Friction Angle (φ)1.00 1.020 10.40 17.50 322.88 12.637 128.86 42.60 466.25 39.792 405.76 91.63 4810.88 3.863 39.39 117.21 3513.25 5.014 51.13 138.00 3614.13 4.081 41.61 145.74 3415.38 8.629 87.99 156.93 3819.00 4.373 44.59 188.78 3423.38 6.793 69.26 227.53 3425.13 10.320 105.23 243.81 3626.88 7.857 80.11 260.10 3429.25 8.308 84.72 281.95 34Sensitive Fine Grained (Dry)Sensitive Fine Grained (Wet)Depth (m)Soil LayerSand to Silty SandSandSand to Silty SandSilty Sand to Sandy SiltSand to Silty SandSandSand to Silty SandSilty Sand to Sandy SiltSand to Silty SandSilty Sand to Sandy SiltLiquefaction Assessment using SPT dataDepth Layer σ'z0CNN60ERr/60 (N1)60rd(CSR)eqkPotential for(m) (kPa) Liquefaction1 Sensitive Fine Grained 17.50 1.57 2.57 1.00 4.04 0.992 0.13 YES2 Clay 35.50 1.33 7.39 1.00 9.87 0.985 0.13 YES3 Clay 43.69 1.27 11.76 1.00 14.87 0.977 0.16 YES4 Clay + Sand to Silty Sand 52.13 1.21 23.37 1.00 28.18 0.969 0.17 NO5 Sand to Silty Sand 61.32 1.15 12.66 1.00 14.58 0.962 0.19 YES6 Sand to Silty Sand + Sand 70.89 1.10 23.06 1.00 25.44 0.954 0.19 NO7 Sand 80.58 1.06 17.43 1.00 18.48 0.946 0.20 YES8 Sand to Silty Sand 89.77 1.02 21.37 1.00 21.89 0.939 0.20 NO9 Sand to Silty Sand 98.96 0.99 10.80 1.00 10.71 0.931 0.21 YES10 Sand to Silty Sand 108.15 0.96 11.18 1.00 10.76 0.96 0.22 YES11 Sand to Silty Sand 117.34 0.93 13.09 1.00 12.24 0.93 0.21 YES12 Sand to Silty Sand 126.53 0.91 13.87 1.00 12.61 0.91 0.21 YES13 Sand to Silty Sand 135.72 0.89 12.13 1.00 10.75 0.88 0.21 YES14 Silty Sand to Sandy Silt 144.41 0.87 10.23 1.00 8.85 0.85 0.20 YES15Silty Sand to Sandy Silt + Sand to Silty Sand153.22 0.85 12.11 1.00 10.24 0.83 0.20 YES16 Sand to Silty Sand 162.41 0.83 17.24 1.00 14.24 0.80 0.19 YES17 Sand to Silty Sand 171.60 0.81 18.58 1.00 15.01 0.77 0.19 YES18 Sand to Silty Sand 180.79 0.79 17.02 1.00 13.45 0.75 0.18 NO19 Sand to Silty Sand 189.98 0.77 19.09 1.00 14.77 0.72 0.18 NO20 Sand 199.67 0.76 18.08 1.00 13.68 0.69 0.17 YES21 Sand 209.36 0.74 20.94 1.00 15.52 0.67 0.16 YES22 Sand 219.05 0.73 22.03 1.00 15.99 0.64 0.16 YES23 Sand 228.74 0.71 21.48 1.00 15.28 0.61 0.15 NO24 Sand to Silty Sand + Sand 238.06 0.70 20.30 1.00 14.17 0.59 0.15 NO25 Sand 247.75 0.68 24.14 1.00 16.53 0.56 0.14 NO26 Sand 257.44 0.67 27.23 1.00 18.30 0.53 0.13 NO27 Sand 267.13 0.66 11.58 1.00 7.64 0.51 0.13 YES28 Sand + Sand to Silty Sand 276.69 0.65 22.12 1.00 14.33 0.48 0.12 NO29 Sand to Silty Sand 285.88 0.64 19.84 1.00 12.63 0.45 0.11 NO30 Sand to Silty Sand + Sand 295.45 0.63 19.93 1.00 12.48 0.43 0.11 NO31 Sand + Sand to Silty Sand 304.76 0.62 19.94 1.00 12.27 0.40 0.10 NO32 Sand to Silty Sand 313.95 0.61 15.93 1.00 9.65 0.37 0.09 YESLiquefaction Assessment using SPT dataDepth Layer σ'z0CNN60ERr/60 (N1)60rd(CSR)eqkPotential for(m) (kPa) Liquefaction1.0 Sensitive Fine Grained (Dry) 17.50 1.57 1.18 1.00 1.86 0.992 0.13 YES2.0 Sensitive Fine Grained (Dry) 35.00 1.34 6.67 1.00 8.93 0.985 0.13 NO3.0 Sensitive Fine Grained (Wet) 43.69 1.27 20.91 1.00 26.46 0.977 0.16 NO4.0Sensitive F.G. (Wet) + Sand to Silty Sand52.63 1.20 17.47 1.00 21.01 0.969 0.17 NO5.0Sand to Silty Sand62.32 1.15 13.92 1.00 15.96 0.962 0.18 NO6.0 Sand to Silty Sand 72.01 1.10 14.72 1.00 16.16 0.954 0.19 NO7.0 Sand to Silty Sand 81.70 1.06 60.72 1.00 64.10 0.946 0.20 NO8.0 Sand to Silty Sand 91.39 1.02 27.05 1.00 27.54 0.939 0.20 NO9.0Sand to Silty Sand + Silty Sand to Sandy Silt100.83 0.99 11.56 1.00 11.39 0.931 0.20 YES10.0 Silty Sand to Sandy Silt 109.52 0.96 7.33 1.00 7.02 0.96 0.21 YES11.0 Silty Sand to Sandy Silt 118.21 0.93 11.49 1.00 10.71 0.93 0.21 YES12.0 Silty Sand to Sandy Silt 126.90 0.91 7.91 1.00 7.18 0.91 0.21 YES13.0 Silty Sand to Sandy Silt 135.59 0.89 11.37 1.00 10.08 0.88 0.21 YES14.0Sand to Silty Sand + Silty Sand to Sandy Silt144.53 0.86 13.53 1.00 11.70 0.85 0.20 YES15.0Silty Sand to Sandy Silt + Sand to Silty Sand153.35 0.85 14.09 1.00 11.90 0.83 0.20 YES16.0 Sand to Silty Sand 162.54 0.83 16.66 1.00 13.75 0.80 0.19 YES17.0 Silty Sand to Sandy Silt 171.23 0.81 11.64 1.00 9.41 0.77 0.19 YES18.0 Silty Sand to Sandy Silt 179.92 0.79 12.70 1.00 10.06 0.75 0.18 YES19.0 Silty Sand to Sandy Silt 188.61 0.78 13.27 1.00 10.30 0.72 0.18 YES20.0 Silty Sand to Sandy Silt 197.30 0.76 13.86 1.00 10.55 0.69 0.17 YES21.0 Silty Sand to Sandy Silt 205.99 0.75 13.72 1.00 10.24 0.67 0.17 YES22.0 Silty Sand to Sandy Silt 214.68 0.73 16.19 1.00 11.86 0.64 0.16 YES23.0 Sand to Silty Sand 223.87 0.72 19.77 1.00 14.21 0.61 0.15 NO24.0 Sand to Silty Sand 233.06 0.71 16.09 1.00 11.35 0.59 0.15 NO25.0 Sand to Silty Sand + Sand 242.37 0.69 18.17 1.00 12.58 0.56 0.14 YES26.0 Sand + Sand to Silty Sand 251.81 0.68 20.27 1.00 13.77 0.53 0.13 YES27.0 Sand to Silty Sand 261.00 0.67 19.08 1.00 12.74 0.51 0.13 YES28.0 Sand to Silty Sand 270.19 0.66 17.02 1.00 11.16 0.48 0.12 NO29.0 Sand to Silty Sand + Sand 279.38 0.64 26.90 1.00 17.34 0.45 0.11 NO30.0 Sand 288.57 0.63 16.34 1.00 10.35 0.43 0.11 YESSENSiTIVE FINE G.RAINED: Avero.§e.- Wo.WCoIl-kftt·, ('t-5t5l.t-St62.2.)/3 -:::. 4o/<tO)o Cc -:: o.o\Sslj- (2.bV-JS) -=- O.oo5tt(2,kJ X I.(-o.'-j--35) ~ e:n7& 2-1 b [-{-l~ CLA"(: Av.5. l,J4ier CoA-Iei1t -4-3!3'7S~k, CC ~ O,trl?t.f -::- 0,0152.. A+ Yoil\1- A: CL.Ny 0::: I%,() /t.N/M> CLjY'"1 ((:/ 18',otwIM~ 01;~:::-:::; AS$VfY'lt<, ;. Ofi\ \ Hfi'll (\ 1 /5 ~JM3) (LOM) + (18 ~NJV\'I~ ((,375M) - (q,gKNIVVl)) CO,~7S{V/) 3g );15 ~o-.. - {8 ,D kf~/v'h3 -::: 13/010/1 OA7(,Cf8Lt7J WI O'c-f77VVl . CPT:R I SIrE P~PA~A1io,-.J (CMt'd) CLA'! : \;\J~ k. -= 1,0 A \0-7 C,N1(>ec, =- 1.0 )(10-'1 Mjsec..Cv =. C-30 : 1<) C+c~O ') ~"'" =- '1,8 I<N JVVi > ~~~-L "'± / Z.Ovn SENSITIvE FIN\': I.~~ G:MI~E".9 CWo'-!) 0-::' Il/5}.:NiM], -----~A_~~7YV) _ 5ENSlTlvE F1NE G'AAINSD (vJ~T) '0:::: l&/SW!VVI) AT H1;1\+ A: u~ ~ (\1-5~I'VYI> )(I.SSM) I (J2:f -Cf~o + !Jeri -dult -SfNS\ TIl/\: FINE GMINED: AVj. Wo.1tr-~-te/lt: \rJ::. 533 cia Cc -= oS59332 eo ~ j-3858 SI'NO -r; 5\G1'( SAN!)-. Cz\-feo:: 0--0152 ,------------,----------------~------------Q'fNA-H Ie CoHt'ACTloN -. \ri~ W:=. ISok.f\l we;~lAt hc.'V'h'VJ~ I-{ ~ 5,0 VV1 h IM'I ~ <N-o f _ 0( ~ 5·0)', \0-3 J(V\/~ ~[150000N)(S/O~)J Y2 P 7 [ (IJ5)\los N) (SYVJ) ] 3D ~ D '/ (D'f{\o-t"llG ~v'n CoN\r~c;h'O/l hl!l~ +0 o.J-her 1'Wi'j'v,~()WI~ ~ o:r lelJ.s.t sfnAc~> ,) /O.%VV1e+res o.\.AJ6.-y SITE PREPARATION CALCULATIONS - CPT #1Cv4.61E-08 m²/sec **Block dimensions are 59"x29.5"x29.5"Hdr1.375 m(δc)ult0.47698 mσ'zo38.575 kPaYears TvU (δc)  (m) σ'zf (kPa)Surcharge (kPa)Height of Wall (m)# of Blocks0.05 0.038 22.1% 0.1055 53.89 -36.11 --- ---0.1 0.077 31.3% 0.1492 61.89 -28.11 --- ---0.15 0.115 38.3% 0.1828 68.83 -21.17 --- ---0.2 0.154 44.3% 0.2111 75.28 -14.72 --- ---0.25 0.192 49.5% 0.2360 81.47 -8.53 --- ---0.3 0.231 54.1% 0.2581 87.39 -2.61 --- ---0.35 0.269 58.3% 0.2779 93.05 3.05 0.17 10.4 0.308 62.0% 0.2960 98.51 8.51 0.47 10.45 0.346 65.5% 0.3124 103.76 13.76 0.76 20.5 0.384 68.6% 0.3273 108.78 18.78 1.04 20.55 0.423 71.5% 0.3408 113.55 23.55 1.31 20.6 0.461 74.0% 0.3531 118.08 28.08 1.56 30.65 0.500 76.4% 0.3643 122.34 32.34 1.80 30.7 0.538 78.5% 0.3745 126.36 36.36 2.02 30.75 0.577 80.5% 0.3838 130.13 40.13 2.23 30.8 0.615 82.2% 0.3923 133.65 43.65 2.43 40.85 0.654 83.8% 0.3999 136.94 46.94 2.61 40.9 0.692 85.3% 0.4069 140.00 50.00 2.78 40.95 0.731 86.6% 0.4132 142.84 52.84 2.94 41 0.769 87.8% 0.4190 145.48 55.48 3.08 51.1 0.846 89.9% 0.4290 150.17 60.17 3.34 51.2 0.923 91.7% 0.4373 154.16 64.16 3.56 51.3 1.000 93.1% 0.4442 157.55 67.55 3.75 61.4 1.077 94.3% 0.4498 160.41 70.41 3.91 61.5 1.153 95.3% 0.4545 162.81 72.81 4.04 61.6 1.230 96.1% 0.4584 164.82 74.82 4.16 61.7 1.307 96.8% 0.4616 166.50 76.50 4.25 61.8 1.384 97.3% 0.4643 167.91 77.91 4.33 61.9 1.461 97.8% 0.4665 169.08 79.08 4.39 62 1.538 98.2% 0.4683 170.06 80.06 4.45 62.2 1.692 98.8% 0.4710 171.54 81.54 4.53 72.4 1.845 99.1% 0.4729 172.57 82.57 4.59 72.6 1.999 99.4% 0.4742 173.27 83.27 4.63 72.8 2.153 99.6% 0.4751 173.75 83.75 4.65 73 2.307 99.7% 0.4757 174.09 84.09 4.67 73.2 2.461 99.8% 0.4761 174.31 84.31 4.68 73.4 2.614 99.9% 0.4764 174.47 84.47 4.69 73.6 2.768 99.9% 0.4766 174.57 84.57 4.70 7SITE PREPARATION CALCULATIONS - CPT #2Cv3.07E-08 m²/sec **Block dimensions are 59"x29.5"x29.5"Hdr1.875 m(δc)ult0.53997 mσ'zo32.8125 kPaYears TvU (δc)  (m) σ'zf (kPa)Surcharge (kPa)Height of Wall (m)# of Blocks0.001 0.000 1.9% 0.0101 33.72 -56.28 --- ---0.003 0.001 3.2% 0.0175 34.40 -55.60 --- ---0.005 0.001 4.2% 0.0226 34.88 -55.12 --- ---0.007 0.002 5.0% 0.0267 35.27 -54.73 --- ---0.01 0.003 5.9% 0.0320 35.77 -54.23 --- ---0.02 0.006 8.4% 0.0452 37.07 -52.93 --- ---0.03 0.008 10.2% 0.0553 38.10 -51.90 --- ---0.04 0.011 11.8% 0.0639 38.99 -51.01 --- ---0.05 0.014 13.2% 0.0714 39.79 -50.21 --- ---0.06 0.017 14.5% 0.0783 40.53 -49.47 --- ---0.07 0.019 15.7% 0.0845 41.22 -48.78 --- ---0.08 0.022 16.7% 0.0904 41.87 -48.13 --- ---0.09 0.025 17.8% 0.0959 42.50 -47.50 --- ---0.1 0.028 18.7% 0.1010 43.10 -46.90 --- ---0.15 0.041 22.9% 0.1238 45.82 -44.18 --- ---0.2 0.055 26.5% 0.1429 48.25 -41.75 --- ---1.2 0.330 64.1% 0.3461 83.49 -6.51 --- ---1.25 0.344 65.3% 0.3526 84.96 -5.04 --- ---1.3 0.358 66.5% 0.3588 86.40 -3.60 --- ---1.35 0.371 67.6% 0.3649 87.83 -2.17 --- ---1.4 0.385 68.7% 0.3707 89.22 -0.78 --- ---1.45 0.399 69.7% 0.3764 90.59 0.59 0.03 11.5 0.413 70.7% 0.3818 91.94 1.94 0.11 11.55 0.426 71.7% 0.3871 93.25 3.25 0.18 11.6 0.440 72.6% 0.3922 94.55 4.55 0.25 11.65 0.454 73.5% 0.3971 95.81 5.81 0.32 11.7 0.468 74.4% 0.4019 97.05 7.05 0.39 11.75 0.481 75.3% 0.4065 98.27 8.27 0.46 11.8 0.495 76.1% 0.4110 99.46 9.46 0.53 11.85 0.509 76.9% 0.4153 100.62 10.62 0.59 11.9 0.523 77.7% 0.4194 101.76 11.76 0.65 11.95 0.536 78.4% 0.4234 102.87 12.87 0.71 12 0.550 79.1% 0.4273 103.95 13.95 0.78 22.05 0.564 79.8% 0.4311 105.01 15.01 0.83 22.1 0.578 80.5% 0.4347 106.05 16.05 0.89 22.15 0.591 81.2% 0.4382 107.05 17.05 0.95 22.2 0.605 81.8% 0.4416 108.04 18.04 1.00 22.25 0.619 82.4% 0.4449 109.00 19.00 1.06 22.3 0.633 83.0% 0.4481 109.94 19.94 1.11 22.35 0.646 83.5% 0.4511 110.85 20.85 1.16 22.4 0.660 84.1% 0.4541 111.74 21.74 1.21 22.45 0.674 84.6% 0.4570 112.61 22.61 1.26 22.5 0.688 85.1% 0.4597 113.45 23.45 1.30 22.55 0.701 85.6% 0.4624 114.28 24.28 1.35 22.6 0.715 86.1% 0.4650 115.08 25.08 1.39 22.65 0.729 86.6% 0.4675 115.86 25.86 1.44 22.7 0.743 87.0% 0.4699 116.61 26.61 1.48 22.75 0.756 87.5% 0.4723 117.35 27.35 1.52 32.8 0.770 87.9% 0.4745 118.07 28.07 1.56 32.85 0.784 88.3% 0.4767 118.77 28.77 1.60 32.9 0.798 88.7% 0.4788 119.45 29.45 1.64 32.95 0.811 89.1% 0.4809 120.11 30.11 1.67 33 0.825 89.4% 0.4828 120.75 30.75 1.71 33.05 0.839 89.8% 0.4847 121.37 31.37 1.74 33.1 0.853 90.1% 0.4866 121.97 31.97 1.78 33.15 0.866 90.4% 0.4884 122.56 32.56 1.81 33.2 0.880 90.8% 0.4901 123.13 33.13 1.84 33.25 0.894 91.1% 0.4918 123.69 33.69 1.87 33.3 0.908 91.4% 0.4934 124.23 34.23 1.90 33.35 0.921 91.7% 0.4949 124.75 34.75 1.93 33.4 0.935 91.9% 0.4964 125.26 35.26 1.96 33.45 0.949 92.2% 0.4979 125.75 35.75 1.99 33.5 0.963 92.5% 0.4993 126.23 36.23 2.01 33.55 0.976 92.7% 0.5006 126.69 36.69 2.04 33.6 0.990 93.0% 0.5019 127.14 37.14 2.06 33.65 1.004 93.2% 0.5032 127.57 37.57 2.09 33.7 1.018 93.4% 0.5044 128.00 38.00 2.11 33.75 1.031 93.6% 0.5056 128.41 38.41 2.13 33.8 1.045 93.9% 0.5068 128.80 38.80 2.16 33.85 1.059 94.1% 0.5079 129.19 39.19 2.18 33.9 1.073 94.3% 0.5090 129.56 39.56 2.20 33.95 1.086 94.4% 0.5100 129.93 39.93 2.22 34 1.100 94.6% 0.5110 130.28 40.28 2.24 36 1.650 98.6% 0.5325 138.07 48.07 2.67 46.75 1.856 99.2% 0.5355 139.18 49.18 2.73 46.8 1.870 99.2% 0.5356 139.24 49.24 2.74 46.85 1.884 99.2% 0.5358 139.29 49.29 2.74 46.9 1.898 99.3% 0.5359 139.35 49.35 2.74 46.95 1.911 99.3% 0.5361 139.40 49.40 2.74 47 1.925 99.3% 0.5362 139.45 49.45 2.75 47.05 1.939 99.3% 0.5363 139.49 49.49 2.75 47.1 1.953 99.3% 0.5364 139.54 49.54 2.75 4BEARING CAPACITY OF SHALLOW FOUNDATIONSTerzaghi and Vesic MethodsDate April 1, 2009Identification CPT 1 - SquareInput ResultsUnits of Measurement Terzaghi VesicSI SI or E Bearing Capacityq ult = 753 kPa 1,063 kPaFoundation Information q a = 251 kPa 354 kPaShape SQ SQ, CI, CO, or REB = 2 m Allowable Column LoadL = m P = 1,004 kN 1,418 kND = 1.5 mSoil Informationc = 0 kPaphi = 30 deggamma = 17.5 kN/m^3Dw = 2 mFactor of SafetyF = 3Copyright 2000 by Donald P. CodutoBEARING CAPACITY OF SHALLOW FOUNDATIONSTerzaghi and Vesic MethodsDate April 1, 2009Identification CPT 1 - CircularInput ResultsUnits of Measurement Terzaghi VesicSI SI or E Bearing Capacityq ult = 735 kPa 1,056 kPaFoundation Information q a = 245 kPa 352 kPaShape CI SQ, CI, CO, or REB = 2.5 m Allowable Column LoadL = m P = 1,203 kN 1,728 kND = 1.5 mSoil Informationc = 0 kPaphi = 30 deggamma = 17.5 kN/m^3Dw = 2 mFactor of SafetyF = 3Copyright 2000 by Donald P. CodutoBEARING CAPACITY OF SHALLOW FOUNDATIONSTerzaghi and Vesic MethodsDate April 1, 2009Identification CPT 1 - ContinuousInput ResultsUnits of Measurement Terzaghi VesicSI SI or E Bearing Capacityq ult = 1,018 kPa 1,027 kPaFoundation Information q a = 339 kPa 342 kPaShape CO SQ, CI, CO, or REB = 3 m Allowable Wall LoadL = m P/b = 1,018 kN/m 1,027 kN/mD = 2 mSoil Informationc = 0 kPaphi = 30 deggamma = 17.5 kN/m^3Dw = 2 mFactor of SafetyF = 3Copyright 2000 by Donald P. CodutoBEARING CAPACITY OF SHALLOW FOUNDATIONSTerzaghi and Vesic MethodsDate April 1, 2009Identification CPT 1 - RectangularInput ResultsUnits of Measurement Terzaghi VesicSI SI or E Bearing Capacityq ult = n/a   kPa 837 kPaFoundation Information q a = n/a   kPa 279 kPaShape RE SQ, CI, CO, or REB = 1 m Allowable Column LoadL = 4 m P = #VALUE! kN 1,115 kND = 1.5 mSoil Informationc = 0 kPaphi = 30 deggamma = 17.5 kN/m^3Dw = 2 mFactor of SafetyF = 3Copyright 2000 by Donald P. CodutoBEARING CAPACITY OF SHALLOW FOUNDATIONSTerzaghi and Vesic MethodsDate April 1, 2009Identification CPT 2 - SquareInput ResultsUnits of Measurement Terzaghi VesicSI SI or E Bearing Capacityq ult = 782 kPa 978 kPaFoundation Information q a = 261 kPa 326 kPaShape SQ SQ, CI, CO, or REB = 2 m Allowable Column LoadL = m P = 1,042 kN 1,305 kND = 1 mSoil Informationc = 0 kPaphi = 32 deggamma = 17.5 kN/m^3Dw = 2 mFactor of SafetyF = 3Copyright 2000 by Donald P. CodutoBEARING CAPACITY OF SHALLOW FOUNDATIONSTerzaghi and Vesic MethodsDate April 1, 2009Identification CPT 2 - CircularInput ResultsUnits of Measurement Terzaghi VesicSI SI or E Bearing Capacityq ult = 731 kPa 988 kPaFoundation Information q a = 244 kPa 329 kPaShape CI SQ, CI, CO, or REB = 2.3 m Allowable Column LoadL = m P = 1,012 kN 1,368 kND = 1 mSoil Informationc = 0 kPaphi = 32 deggamma = 17.5 kN/m^3Dw = 2 mFactor of SafetyF = 3Copyright 2000 by Donald P. CodutoBEARING CAPACITY OF SHALLOW FOUNDATIONSTerzaghi and Vesic MethodsDate April 1, 2009Identification CPT 2 - ContinuousInput ResultsUnits of Measurement Terzaghi VesicSI SI or E Bearing Capacityq ult = 1,268 kPa 1,281 kPaFoundation Information q a = 423 kPa 427 kPaShape CO SQ, CI, CO, or REB = 2.5 m Allowable Wall LoadL = m P/b = 1,057 kN/m 1,068 kN/mD = 2 mSoil Informationc = 0 kPaphi = 32 deggamma = 17.5 kN/m^3Dw = 2 mFactor of SafetyF = 3Copyright 2000 by Donald P. CodutoBEARING CAPACITY OF SHALLOW FOUNDATIONSTerzaghi and Vesic MethodsDate April 1, 2009Identification CPT 2 - RectangularInput ResultsUnits of Measurement Terzaghi VesicSI SI or E Bearing Capacityq ult = n/a   kPa 1,422 kPaFoundation Information q a = n/a   kPa 474 kPaShape RE SQ, CI, CO, or REB = 1 m Allowable Column LoadL = 2.5 m P = #VALUE! kN 1,185 kND = 2 mSoil Informationc = 0 kPaphi = 32 deggamma = 17.5 kN/m^3Dw = 2 mFactor of SafetyF = 3Copyright 2000 by Donald P. CodutoSETTLEMENT ANALYSIS OF SHALLOW FOUNDATIONSClassical MethodDate March 28, 2009Identification CPT 1 - SquareInput ResultsUnits SI E or SIShape SQ SQ, CI, CO, or RE q = 390 kPaB = 2 m delta = 308.09 mmL = mD = 1.5 mP = 1418 kNDw = 2 mr = 0.85  Depth to Soil LayerTop Bottom Cc/(1+e) Cr/(1+e) sigma m' gamma zf sigma c' sigma zo' delta sigma sigma zf' strain delta(m) (m) (kPa) (kN/m^3) (m) (kPa) (kPa) (kPa) (kPa) (%) (mm)0.0 1.5 17.51.5 1.6 0.197 0.066 0 18 0.05 27 27 364 391 19.39 19.3931.6 1.7 0.197 0.066 0 18 0.15 29 29 363 392 18.94 18.9441.7 1.8 0.197 0.066 0 18 0.25 31 31 360 390 18.48 18.4821.8 1.9 0.197 0.066 0 18 0.35 33 33 354 386 17.99 17.9871.9 2.0 0.197 0.066 0 18 0.45 34 34 344 378 17.45 17.4492.0 2.1 0.197 0.066 0 18 0.55 36 36 331 367 16.96 16.9572.1 2.2 0.197 0.066 0 18 0.65 36 36 316 353 16.50 16.5042.2 2.3 0.197 0.066 0 18 0.75 37 37 300 337 16.01 16.0052.3 2.4 0.197 0.066 0 18 0.85 38 38 282 320 15.47 15.4722.4 2.5 0.197 0.066 0 18 0.95 39 39 264 303 14.92 14.9152.5 2.6 0.197 0.066 0 18 1.05 40 40 246 286 14.34 14.3442.6 2.7 0.197 0.066 0 18 1.15 41 41 229 269 13.77 13.7662.7 2.8 0.197 0.066 0 18 1.25 41 41 212 254 13.19 13.1892.8 2.9 0.197 0.066 0 18 1.35 42 42 197 239 12.62 12.6172.9 3.0 0.197 0.066 0 18 1.45 43 43 183 226 12.06 12.0563.0 3.1 0.197 0.066 0 18 1.55 44 44 170 213 11.51 11.5073.1 3.2 0.197 0.066 0 18 1.65 45 45 157 202 10.98 10.9753.2 3.3 0.197 0.066 0 18 1.75 45 45 146 192 10.46 10.4603.3 3.4 0.197 0.066 0 18 1.85 46 46 136 182 9.96 9.9643.4 3.5 0.197 0.066 0 18 1.95 47 47 127 174 9.49 9.4873.5 3.6 0.197 0.066 0 18 2.05 48 48 118 166 9.03 9.0293.6 3.7 0.197 0.066 0 18 2.15 49 49 110 159 8.59 8.592Copyright 2000 by Donald P. CodutoSETTLEMENT ANALYSIS OF SHALLOW FOUNDATIONSClassical MethodDate March 28, 2009Identification CPT 1 - CircularInput ResultsUnits SI E or SIShape CI SQ, CI, CO, or RE q = 388 kPaB = 2.5 m delta = 305.46 mmL = mD = 1.5 mP = 1728 kNDw = 2 mr = 0.85  Depth to Soil LayerTop Bottom Cc/(1+e) Cr/(1+e) sigma m' gamma zf sigma c' sigma zo' delta sigma sigma zf' strain delta(m) (m) (kPa) (kN/m^3) (m) (kPa) (kPa) (kPa) (kPa) (%) (mm)0.0 1.5 17.51.5 1.6 0.197 0.066 0 18 0.05 27 27 361 388 19.35 19.3501.6 1.7 0.197 0.066 0 18 0.15 29 29 360 389 18.90 18.9001.7 1.8 0.197 0.066 0 18 0.25 31 31 357 388 18.44 18.4351.8 1.9 0.197 0.066 0 18 0.35 33 33 351 383 17.94 17.9351.9 2.0 0.197 0.066 0 18 0.45 34 34 341 375 17.39 17.3892.0 2.1 0.197 0.066 0 18 0.55 36 36 328 364 16.89 16.8862.1 2.2 0.197 0.066 0 18 0.65 36 36 312 349 16.42 16.4212.2 2.3 0.197 0.066 0 18 0.75 37 37 295 333 15.91 15.9102.3 2.4 0.197 0.066 0 18 0.85 38 38 277 315 15.36 15.3652.4 2.5 0.197 0.066 0 18 0.95 39 39 259 298 14.80 14.7962.5 2.6 0.197 0.066 0 18 1.05 40 40 241 281 14.21 14.2142.6 2.7 0.197 0.066 0 18 1.15 41 41 224 264 13.63 13.6272.7 2.8 0.197 0.066 0 18 1.25 41 41 207 249 13.04 13.0422.8 2.9 0.197 0.066 0 18 1.35 42 42 192 234 12.46 12.4632.9 3.0 0.197 0.066 0 18 1.45 43 43 178 221 11.90 11.8963.0 3.1 0.197 0.066 0 18 1.55 44 44 165 209 11.34 11.3433.1 3.2 0.197 0.066 0 18 1.65 45 45 153 197 10.81 10.8083.2 3.3 0.197 0.066 0 18 1.75 45 45 142 187 10.29 10.2903.3 3.4 0.197 0.066 0 18 1.85 46 46 132 178 9.79 9.7923.4 3.5 0.197 0.066 0 18 1.95 47 47 123 170 9.31 9.3143.5 3.6 0.197 0.066 0 18 2.05 48 48 114 162 8.86 8.8573.6 3.7 0.197 0.066 0 18 2.15 49 49 106 155 8.42 8.420Copyright 2000 by Donald P. CodutoSETTLEMENT ANALYSIS OF SHALLOW FOUNDATIONSClassical MethodDate March 28, 2009Identification CPT 1 - ContinuousInput ResultsUnits SI E or SIShape CO SQ, CI, CO, or RE q = 390 kPaB = 3 m delta = 269.34 mmL = mD = 2 mP = 1027 kN/mDw = 2 mr = 0.85  Depth to Soil LayerTop Bottom Cc/(1+e) Cr/(1+e) sigma m' gamma zf sigma c' sigma zo' delta sigma sigma zf' strain delta(m) (m) (kPa) (kN/m^3) (m) (kPa) (kPa) (kPa) (kPa) (%) (mm)0.0 2.0 17.52.0 2.1 0.197 0.066 0 18 0.05 35 35 355 390 17.45 17.4462.1 2.2 0.197 0.066 0 18 0.15 36 36 354 391 17.29 17.2922.2 2.3 0.197 0.066 0 18 0.25 37 37 354 391 17.13 17.1352.3 2.4 0.197 0.066 0 18 0.35 38 38 353 391 16.97 16.9702.4 2.5 0.197 0.066 0 18 0.45 39 39 351 390 16.79 16.7952.5 2.6 0.197 0.066 0 18 0.55 40 40 348 388 16.61 16.6072.6 2.7 0.197 0.066 0 18 0.65 40 40 345 385 16.41 16.4062.7 2.8 0.197 0.066 0 18 0.75 41 41 340 381 16.19 16.1912.8 2.9 0.197 0.066 0 18 0.85 42 42 335 377 15.96 15.9642.9 3.0 0.197 0.066 0 18 0.95 43 43 329 372 15.73 15.7253.0 3.1 0.197 0.066 0 18 1.05 44 44 323 366 15.48 15.4773.1 3.2 0.197 0.066 0 18 1.15 44 44 316 360 15.22 15.2213.2 3.3 0.197 0.066 0 18 1.25 45 45 309 354 14.96 14.9593.3 3.4 0.197 0.066 0 18 1.35 46 46 301 347 14.69 14.6933.4 3.5 0.197 0.066 0 18 1.45 47 47 294 341 14.42 14.4243.5 3.6 0.197 0.066 0 18 1.55 48 48 286 334 14.15 14.1533.6 3.7 0.197 0.066 0 18 1.65 49 49 279 327 13.88 13.883Copyright 2000 by Donald P. CodutoSETTLEMENT ANALYSIS OF SHALLOW FOUNDATIONSClassical MethodDate March 28, 2009Identification CPT 1 - RectangularInput ResultsUnits SI E or SIShape RE SQ, CI, CO, or RE q = 314 kPaB = 1 m delta = 253.87 mmL = 4 mD = 1.5 mP = 1115 kNDw = 2 mr = 0.85  Depth to Soil LayerTop Bottom Cc/(1+e) Cr/(1+e) sigma m' gamma zf sigma c' sigma zo' delta sigma sigma zf' strain delta(m) (m) (kPa) (kN/m^3) (m) (kPa) (kPa) (kPa) (kPa) (%) (mm)0.0 1.5 17.51.5 1.6 0.197 0.066 0 18 0.05 27 27 288 315 17.82 17.8241.6 1.7 0.197 0.066 0 18 0.15 29 29 285 314 17.33 17.3331.7 1.8 0.197 0.066 0 18 0.25 31 31 276 307 16.73 16.7311.8 1.9 0.197 0.066 0 18 0.35 33 33 262 294 16.02 16.0161.9 2.0 0.197 0.066 0 18 0.45 34 34 244 279 15.23 15.2252.0 2.1 0.197 0.066 0 18 0.55 36 36 226 262 14.49 14.4892.1 2.2 0.197 0.066 0 18 0.65 36 36 208 244 13.83 13.8282.2 2.3 0.197 0.066 0 18 0.75 37 37 191 228 13.17 13.1732.3 2.4 0.197 0.066 0 18 0.85 38 38 176 214 12.54 12.5372.4 2.5 0.197 0.066 0 18 0.95 39 39 162 201 11.93 11.9262.5 2.6 0.197 0.066 0 18 1.05 40 40 149 189 11.34 11.3422.6 2.7 0.197 0.066 0 18 1.15 41 41 138 179 10.79 10.7862.7 2.8 0.197 0.066 0 18 1.25 41 41 128 170 10.26 10.2572.8 2.9 0.197 0.066 0 18 1.35 42 42 119 161 9.75 9.7532.9 3.0 0.197 0.066 0 18 1.45 43 43 111 154 9.28 9.2753.0 3.1 0.197 0.066 0 18 1.55 44 44 104 148 8.82 8.8213.1 3.2 0.197 0.066 0 18 1.65 45 45 97 142 8.39 8.3883.2 3.3 0.197 0.066 0 18 1.75 45 45 91 136 7.98 7.9773.3 3.4 0.197 0.066 0 18 1.85 46 46 85 131 7.59 7.5863.4 3.5 0.197 0.066 0 18 1.95 47 47 80 127 7.21 7.2153.5 3.6 0.197 0.066 0 18 2.05 48 48 75 123 6.86 6.8623.6 3.7 0.197 0.066 0 18 2.15 49 49 71 120 6.53 6.526Copyright 2000 by Donald P. CodutoSETTLEMENT ANALYSIS OF SHALLOW FOUNDATIONSClassical MethodDate March 28, 2009Identification CPT 2 - SquareInput ResultsUnits SI E or SIShape SQ SQ, CI, CO, or RE q = 350 kPaB = 2 m delta = 396.76 mmL = mD = 1 mP = 1305 kNDw = 2 mr = 0.85  Depth to Soil LayerTop Bottom Cc/(1+e) Cr/(1+e) sigma m' gamma zf sigma c' sigma zo' delta sigma sigma zf' strain delta(m) (m) (kPa) (kN/m^3) (m) (kPa) (kPa) (kPa) (kPa) (%) (mm)0.0 1.0 17.51.0 1.1 0.22755 0.07585 0 17.5 0.05 18 18 332 351 24.77 24.7711.1 1.2 0.22755 0.07585 0 17.5 0.15 20 20 332 352 24.03 24.0301.2 1.3 0.22755 0.07585 0 17.5 0.25 22 22 329 351 23.30 23.3051.3 1.4 0.22755 0.07585 0 17.5 0.35 24 24 323 347 22.57 22.5651.4 1.5 0.22755 0.07585 0 17.5 0.45 25 25 314 340 21.80 21.7961.5 1.6 0.22755 0.07585 0 17.5 0.55 27 27 303 330 20.99 20.9911.6 1.7 0.22755 0.07585 0 17.5 0.65 29 29 289 318 20.15 20.1541.7 1.8 0.22755 0.07585 0 17.5 0.75 31 31 274 304 19.29 19.2921.8 1.9 0.22755 0.07585 0 17.5 0.85 32 32 258 290 18.42 18.4171.9 2.0 0.22755 0.07585 0 17.5 0.95 34 34 241 275 17.54 17.5362.0 2.1 0.22755 0.07585 0 18.5 1.05 35 35 225 260 16.75 16.7502.1 2.2 0.22755 0.07585 0 18.5 1.15 36 36 209 245 16.05 16.0532.2 2.3 0.22755 0.07585 0 18.5 1.25 37 37 194 231 15.36 15.3582.3 2.4 0.22755 0.07585 0 18.5 1.35 38 38 180 218 14.67 14.6722.4 2.5 0.22755 0.07585 0 18.5 1.45 39 39 167 206 14.00 13.9982.5 2.6 0.22755 0.07585 0 18.5 1.55 40 40 155 195 13.34 13.3422.6 2.7 0.22755 0.07585 0 18.5 1.65 41 41 144 185 12.71 12.7062.7 2.8 0.22755 0.07585 0 18.5 1.75 42 42 134 175 12.09 12.0912.8 2.9 0.22755 0.07585 0 18.5 1.85 42 42 124 167 11.50 11.5002.9 3.0 0.22755 0.07585 0 18.5 1.95 43 43 116 159 10.93 10.9323.0 3.1 0.22755 0.07585 0 18.5 2.05 44 44 108 152 10.39 10.3883.1 3.2 0.22755 0.07585 0 18.5 2.15 45 45 101 146 9.87 9.8693.2 3.3 0.22755 0.07585 0 18.5 2.25 46 46 94 140 9.37 9.3743.3 3.4 0.22755 0.07585 0 18.5 2.35 47 47 88 135 8.90 8.9033.4 3.5 0.0752 0.025067 0 18.5 2.45 48 48 83 130 2.79 2.7943.5 3.6 0.0752 0.025067 0 18.5 2.55 48 48 78 126 2.65 2.6533.6 3.7 0.0752 0.025067 0 18.5 2.65 49 49 73 122 2.52 2.520Copyright 2000 by Donald P. CodutoSETTLEMENT ANALYSIS OF SHALLOW FOUNDATIONSClassical MethodDate March 28, 2009Identification CPT 2 - CircularInput ResultsUnits SI E or SIShape CI SQ, CI, CO, or RE q = 353 kPaB = 2.3 m delta = 398.83 mmL = mD = 1 mP = 1368 kNDw = 2 mr = 0.85  Depth to Soil LayerTop Bottom Cc/(1+e) Cr/(1+e) sigma m' gamma zf sigma c' sigma zo' delta sigma sigma zf' strain delta(m) (m) (kPa) (kN/m^3) (m) (kPa) (kPa) (kPa) (kPa) (%) (mm)0.0 1.0 17.51.0 1.1 0.22755 0.07585 0 17.5 0.05 18 18 335 354 24.85 24.8471.1 1.2 0.22755 0.07585 0 17.5 0.15 20 20 334 355 24.10 24.0981.2 1.3 0.22755 0.07585 0 17.5 0.25 22 22 331 353 23.35 23.3511.3 1.4 0.22755 0.07585 0 17.5 0.35 24 24 323 347 22.57 22.5691.4 1.5 0.22755 0.07585 0 17.5 0.45 25 25 312 338 21.74 21.7391.5 1.6 0.22755 0.07585 0 17.5 0.55 27 27 298 325 20.86 20.8601.6 1.7 0.22755 0.07585 0 17.5 0.65 29 29 281 310 19.94 19.9411.7 1.8 0.22755 0.07585 0 17.5 0.75 31 31 263 294 19.00 18.9951.8 1.9 0.22755 0.07585 0 17.5 0.85 32 32 245 277 18.04 18.0381.9 2.0 0.22755 0.07585 0 17.5 0.95 34 34 227 261 17.08 17.0822.0 2.1 0.22755 0.07585 0 18.5 1.05 35 35 209 245 16.23 16.2262.1 2.2 0.22755 0.07585 0 18.5 1.15 36 36 193 229 15.47 15.4662.2 2.3 0.22755 0.07585 0 18.5 1.25 37 37 177 214 14.72 14.7172.3 2.4 0.22755 0.07585 0 18.5 1.35 38 38 163 201 13.98 13.9842.4 2.5 0.22755 0.07585 0 18.5 1.45 39 39 150 189 13.27 13.2732.5 2.6 0.22755 0.07585 0 18.5 1.55 40 40 138 178 12.59 12.5862.6 2.7 0.22755 0.07585 0 18.5 1.65 41 41 127 168 11.93 11.9262.7 2.8 0.22755 0.07585 0 18.5 1.75 42 42 118 159 11.29 11.2942.8 2.9 0.22755 0.07585 0 18.5 1.85 42 42 109 151 10.69 10.6902.9 3.0 0.22755 0.07585 0 18.5 1.95 43 43 101 144 10.12 10.1153.0 3.1 0.22755 0.07585 0 18.5 2.05 44 44 94 138 9.57 9.5693.1 3.2 0.22755 0.07585 0 18.5 2.15 45 45 87 132 9.05 9.0513.2 3.3 0.22755 0.07585 0 18.5 2.25 46 46 81 127 8.56 8.5613.3 3.4 0.22755 0.07585 0 18.5 2.35 47 47 76 123 8.10 8.0973.4 3.5 0.22755 0.07585 0 18.5 2.45 48 48 71 118 7.66 7.6593.5 3.6 0.22755 0.07585 0 18.5 2.55 48 48 66 115 7.25 7.2453.6 3.7 0.22755 0.07585 0 18.5 2.65 49 49 62 112 6.86 6.855Copyright 2000 by Donald P. CodutoSETTLEMENT ANALYSIS OF SHALLOW FOUNDATIONSClassical MethodDate March 28, 2009Identification CPT 2 - ContinuousInput ResultsUnits SI E or SIShape CO SQ, CI, CO, or RE q = 474 kPaB = 2.5 m delta = 332.73 mmL = mD = 2 mP = 1068 kN/mDw = 2 mr = 0.85  Depth to Soil LayerTop Bottom Cc/(1+e) Cr/(1+e) sigma m' gamma zf sigma c' sigma zo' delta sigma sigma zf' strain delta(m) (m) (kPa) (kN/m^3) (m) (kPa) (kPa) (kPa) (kPa) (%) (mm)0.0 2.0 17.52.0 2.1 0.22755 0.07585 0 18.5 0.05 35 35 439 475 21.80 21.8002.1 2.2 0.22755 0.07585 0 18.5 0.15 36 36 439 475 21.61 21.6062.2 2.3 0.22755 0.07585 0 18.5 0.25 37 37 438 475 21.40 21.4032.3 2.4 0.22755 0.07585 0 18.5 0.35 38 38 436 474 21.18 21.1832.4 2.5 0.22755 0.07585 0 18.5 0.45 39 39 432 471 20.94 20.9422.5 2.6 0.22755 0.07585 0 18.5 0.55 40 40 427 466 20.68 20.6762.6 2.7 0.22755 0.07585 0 18.5 0.65 41 41 420 460 20.39 20.3872.7 2.8 0.22755 0.07585 0 18.5 0.75 42 42 412 453 20.08 20.0762.8 2.9 0.22755 0.07585 0 18.5 0.85 42 42 402 445 19.75 19.7452.9 3.0 0.22755 0.07585 0 18.5 0.95 43 43 392 436 19.40 19.4003.0 3.1 0.22755 0.07585 0 18.5 1.05 44 44 382 426 19.04 19.0423.1 3.2 0.22755 0.07585 0 18.5 1.15 45 45 371 416 18.68 18.6763.2 3.3 0.22755 0.07585 0 18.5 1.25 46 46 360 405 18.30 18.3043.3 3.4 0.22755 0.07585 0 18.5 1.35 47 47 348 395 17.93 17.9303.4 3.5 0.22755 0.07585 0 18.5 1.45 48 48 337 385 17.56 17.5563.5 3.6 0.22755 0.07585 0 18.5 1.55 48 48 327 375 17.18 17.1843.6 3.7 0.22755 0.07585 0 18.5 1.65 49 49 316 365 16.82 16.815Copyright 2000 by Donald P. CodutoSETTLEMENT ANALYSIS OF SHALLOW FOUNDATIONSClassical MethodDate March 28, 2009Identification CPT 2 - RectangularInput ResultsUnits SI E or SIShape RE SQ, CI, CO, or RE q = 521 kPaB = 1 m delta = 292.13 mmL = 2.5 mD = 2 mP = 1185 kNDw = 2 mr = 0.85  Depth to Soil LayerTop Bottom Cc/(1+e) Cr/(1+e) sigma m' gamma zf sigma c' sigma zo' delta sigma sigma zf' strain delta(m) (m) (kPa) (kN/m^3) (m) (kPa) (kPa) (kPa) (kPa) (%) (mm)0.0 2.0 17.52.0 2.1 0.22755 0.07585 0 18.5 0.05 35 35 486 521 22.59 22.5872.1 2.2 0.22755 0.07585 0 18.5 0.15 36 36 481 517 22.32 22.3172.2 2.3 0.22755 0.07585 0 18.5 0.25 37 37 466 503 21.88 21.8822.3 2.4 0.22755 0.07585 0 18.5 0.35 38 38 441 479 21.28 21.2762.4 2.5 0.22755 0.07585 0 18.5 0.45 39 39 410 449 20.54 20.5422.5 2.6 0.22755 0.07585 0 18.5 0.55 40 40 377 417 19.73 19.7312.6 2.7 0.22755 0.07585 0 18.5 0.65 41 41 344 385 18.88 18.8852.7 2.8 0.22755 0.07585 0 18.5 0.75 42 42 314 355 18.03 18.0282.8 2.9 0.22755 0.07585 0 18.5 0.85 42 42 285 328 17.18 17.1792.9 3.0 0.22755 0.07585 0 18.5 0.95 43 43 260 303 16.35 16.3463.0 3.1 0.22755 0.07585 0 18.5 1.05 44 44 236 281 15.54 15.5373.1 3.2 0.22755 0.07585 0 18.5 1.15 45 45 216 261 14.75 14.7543.2 3.3 0.22755 0.07585 0 18.5 1.25 46 46 197 243 14.00 13.9993.3 3.4 0.22755 0.07585 0 18.5 1.35 47 47 180 227 13.28 13.2763.4 3.5 0.22755 0.07585 0 18.5 1.45 48 48 165 213 12.58 12.5823.5 3.6 0.22755 0.07585 0 18.5 1.55 48 48 152 200 11.92 11.9213.6 3.7 0.22755 0.07585 0 18.5 1.65 49 49 140 189 11.29 11.290Copyright 2000 by Donald P. Codutokv120 MPa/m (Page 129 CFEM)q 19.152 kPa 400psf Max Live Loadsδ 0.000958 m0.9576 mmkvb0.109718 MPa/mq 19.152 kPaδ 0.174557 m174.5566 mmI 1 (page 161 CFEM)q 19.152 kPav 0.3E 4.788026 MPa http://www.geotechnicalinfo.com/youngs_modulus.html 50tsfb 0.263078 mRAFT FOUNDATION PROPERTIESkv18 MPa/m (Page 129 CFEM)q 19.152 kPa 400psf Max Live Loadsδ 0.002394 m2.394 mmkvb0.793479 MPa/mq 19.152 kPaδ 0.024137 m24.13674 mmI 1 (page 161 CFEM)q 19.152 kPav 0.3E 4.788026 MPa http://www.geotechnicalinfo.com/youngs_modulus.html 50tsfb 0.657696 mRAFT BEARING ANALYSISBEARING CAPACITY OF SHALLOW FOUNDATIONSTerzaghi and Vesic MethodsDate March 28, 2009Identification CPT 1Input ResultsUnits of Measurement Terzaghi VesicSI SI or E Bearing Capacityq ult = 243 kPa 263 kPaFoundation Information q a = 81 kPa 88 kPaShape CO SQ, CI, CO, or REB = 0.263078 m Allowable Wall LoadL = m P/b = 21 kN/m 23 kN/mD = 0.5 mSoil Informationc = 0 kPaphi = 30 deggamma = 17.5 kN/m^3Dw = 2 mFactor of SafetyF = 3Copyright 2000 by Donald P. CodutoRAFT BEARING ANALYSISBEARING CAPACITY OF SHALLOW FOUNDATIONSTerzaghi and Vesic MethodsDate March 28, 2009Identification CPT 2Input ResultsUnits of Measurement Terzaghi VesicSI SI or E Bearing Capacityq ult = 312 kPa 325 kPaFoundation Information q a = 104 kPa 108 kPaShape CO SQ, CI, CO, or REB = 0.657696 m Allowable Wall LoadL = m P/b = 68 kN/m 71 kN/mD = 0.5 mSoil Informationc = 0 kPaphi = 30 deggamma = 17.5 kN/m^3Dw = 2 mFactor of SafetyF = 3Copyright 2000 by Donald P. CodutoSETTLEMENT ANALYSIS OF SHALLOW FOUNDATIONSClassical MethodDate March 28, 2009Identification CPT 1 - RaftInput ResultsUnits SI E or SIShape CO SQ, CI, CO, or RE q = 92 kPaB = 0.263078 m delta = 106.83 mmL = mD = 0.5 mP = 21 kN/mDw = 2 mr = 0.85  Depth to Soil LayerTop Bottom Cc/(1+e) Cr/(1+e) sigma m' gamma zf sigma c' sigma zo' delta sigma sigma zf' strain delta(m) (m) (kPa) (kN/m^3) (m) (kPa) (kPa) (kPa) (kPa) (%) (mm)0.0 0.5 17.50.5 0.6 0.197 0.065667 0 17.5 0.05 10 10 81 91 16.33 16.3260.6 0.7 0.197 0.065667 0 17.5 0.15 11 11 64 76 13.77 13.7650.7 0.8 0.197 0.065667 0 17.5 0.25 13 13 47 60 11.10 11.1030.8 0.9 0.197 0.065667 0 17.5 0.35 15 15 36 51 8.99 8.9910.9 1.0 0.197 0.065667 0 17.5 0.45 17 17 29 46 7.38 7.3751.0 1.1 0.197 0.065667 0 18 0.55 18 18 24 43 6.13 6.1261.1 1.2 0.197 0.065667 0 18 0.65 20 20 21 41 5.15 5.1461.2 1.3 0.197 0.065667 0 18 0.75 22 22 18 40 4.37 4.3731.3 1.4 0.197 0.065667 0 18 0.85 24 24 16 40 3.75 3.7531.4 1.5 0.197 0.065667 0 18 0.95 26 26 14 40 3.25 3.2501.5 1.6 0.197 0.065667 0 18 1.05 27 27 13 40 2.84 2.8391.6 1.7 0.197 0.065667 0 18 1.15 29 29 12 41 2.50 2.4981.7 1.8 0.197 0.065667 0 18 1.25 31 31 11 42 2.21 2.2121.8 1.9 0.197 0.065667 0 18 1.35 33 33 10 43 1.97 1.9721.9 2.0 0.197 0.065667 0 18 1.45 35 35 10 44 1.77 1.7682.0 2.1 0.197 0.065667 0 18 1.55 36 36 9 45 1.61 1.6122.1 2.2 0.197 0.065667 0 18 1.65 37 37 8 45 1.49 1.4942.2 2.3 0.197 0.065667 0 18 1.75 38 38 8 46 1.23 1.2252.3 2.4 0.197 0.065667 0 18 1.85 39 39 7 46 1.14 1.1402.4 2.5 0.197 0.065667 0 18 1.95 39 39 7 47 1.06 1.0632.5 2.6 0.197 0.065667 0 18 2.05 40 40 7 47 0.99 0.9942.6 2.7 0.197 0.065667 0 18 2.15 41 41 6 48 0.93 0.9312.7 2.8 0.197 0.065667 0 18 2.25 42 42 6 48 0.87 0.8742.8 2.9 0.197 0.065667 0 18 2.35 43 43 6 49 0.82 0.8232.9 3.0 0.197 0.065667 0 18 2.45 44 44 6 50 0.78 0.7763.0 3.1 0.197 0.065667 0 18 2.55 45 45 5 50 0.73 0.7333.1 3.2 0.197 0.065667 0 18 2.65 46 46 5 51 0.69 0.6933.2 3.3 0.197 0.065667 0 18 2.75 47 47 5 52 0.66 0.6573.3 3.4 0.197 0.065667 0 18 2.85 48 48 5 53 0.62 0.6233.4 3.5 0.197 0.065667 0 18 2.95 49 49 5 53 0.59 0.5923.5 3.6 0.197 0.065667 0 18 3.05 50 50 5 54 0.56 0.5643.6 3.7 0.197 0.065667 0 18 3.15 50 50 4 55 0.54 0.537Copyright 2000 by Donald P. CodutoSETTLEMENT ANALYSIS OF SHALLOW FOUNDATIONSClassical MethodDate March 28, 2009Identification CPT 2 - RaftInput ResultsUnits SI E or SIShape CO SQ, CI, CO, or RE q = 115.191 kPaB = 0.657696 m delta = 192.62 mmL = mD = 0.5 mP = 68 kN/mDw = 2 mr = 0.85  Depth to Soil LayerTop Bottom Cc/(1+e) Cr/(1+e) sigma m' gamma zf sigma c' sigma zo' delta sigma sigma zf' strain delta(m) (m) (kPa) (kN/m^3) (m) (kPa) (kPa) (kPa) (kPa) (%) (mm)0.0 0.5 17.50.5 0.6 0.22755 0.07585 0 17.5 0.05 10 10 106 116 20.90 20.9030.6 0.7 0.22755 0.07585 0 17.5 0.15 11 11 103 114 19.39 19.3900.7 0.8 0.22755 0.07585 0 17.5 0.25 13 13 95 108 17.72 17.7170.8 0.9 0.22755 0.07585 0 17.5 0.35 15 15 85 100 15.99 15.9890.9 1.0 0.22755 0.07585 0 17.5 0.45 17 17 75 92 14.34 14.3411.0 1.1 0.22755 0.07585 0 17.5 0.55 18 18 66 85 12.82 12.8221.1 1.2 0.22755 0.07585 0 17.5 0.65 20 20 59 79 11.46 11.4571.2 1.3 0.22755 0.07585 0 17.5 0.75 22 22 53 75 10.26 10.2601.3 1.4 0.22755 0.07585 0 17.5 0.85 24 24 48 72 9.21 9.2131.4 1.5 0.22755 0.07585 0 17.5 0.95 26 26 44 69 8.30 8.2981.5 1.6 0.22755 0.07585 0 17.5 1.05 28 28 40 68 7.50 7.4981.6 1.7 0.22755 0.07585 0 17.5 1.15 30 30 37 66 6.80 6.7971.7 1.8 0.22755 0.07585 0 17.5 1.25 31 31 34 65 6.18 6.1801.8 1.9 0.22755 0.07585 0 17.5 1.35 33 33 32 65 5.64 5.6371.9 2.0 0.22755 0.07585 0 17.5 1.45 35 35 30 65 5.16 5.1572.0 2.1 0.22755 0.07585 0 18.5 1.55 36 36 28 64 4.78 4.7792.1 2.2 0.22755 0.07585 0 18.5 1.65 37 37 26 64 4.48 4.4852.2 2.3 0.22755 0.07585 0 18.5 1.75 38 38 25 63 4.22 4.2162.3 2.4 0.22755 0.07585 0 18.5 1.85 39 39 24 63 3.97 3.9712.4 2.5 0.22755 0.07585 0 18.5 1.95 40 40 22 62 1.24 1.2382.5 2.6 0.22755 0.07585 0 18.5 2.05 41 41 21 62 1.17 1.1702.6 2.7 0.22755 0.07585 0 18.5 2.15 42 42 20 62 1.11 1.1072.7 2.8 0.22755 0.07585 0 18.5 2.25 42 42 20 61 0.00 0.0002.8 2.9 0.22755 0.07585 0 18.5 2.35 41 41 19 59 0.00 0.0002.9 3.0 0.22755 0.07585 0 18.5 2.45 40 40 18 58 0.00 0.0003.0 3.1 0.22755 0.07585 0 18.5 2.55 39 39 17 56 0.00 0.0003.1 3.2 0.22755 0.07585 0 18.5 2.65 38 38 17 54 0.00 0.0003.2 3.3 0.22755 0.07585 0 18.5 2.75 37 37 16 53 0.00 0.0003.3 3.4 0.22755 0.07585 0 18.5 2.85 36 36 16 51 0.00 0.0003.4 3.5 0.0752 0.025067 0 18.5 2.95 35 35 15 50 0.00 0.0003.5 3.6 0.0752 0.025067 0 18.5 3.05 34 34 15 48 0.00 0.0003.6 3.7 0.0752 0.025067 0 18.5 3.15 33 33 14 47 0.00 0.000Copyright 2000 by Donald P. CodutoProject Number: Title: Squamish Shoppers Drug Mart Date: Parish: Station:      Offset: Elevation (ft): Pile Shape: End Bearing  & Ultimate CapacityRound CPT No: 1Pile Diameter: 1001020304050607080901000 10 20 30 40 50 60 70 80 90 100110120130140150160170180190200210Depth (ft)Predicted Pile Capacity (tons)Project Number: Title: Date: Parish: Station:      Offset: Elevation (ft): Pile Shape:  End Bearing & Ultimate CapacityRound CPT No: Pile Diameter: 1201020304050607080901000 20 40 60 80 100 120 140 160 180 200 220 240 260Depth (ft)Predicted Pile Capacity (tons)Project Number: Title: Squamish Shoppers Drug Mart Date: Parish: Station:      Offset: Elevation (ft): Pile Shape: End Bearing  & Ultimate CapacityRound CPT No: 1Pile Diameter: 1601020304050607080901000 20 40 60 80 100 120 140 160 180 200 220 240 260 280 300 320 340 360Depth (ft)Predicted Pile Capacity (tons)Project Number: Title: Squamish Shoppers Drug Mart Date: Parish: Station:      Offset: Elevation (ft): Pile Shape: End Bearing  & Ultimate CapacityRound CPT No: 2Pile Diameter: 1001020304050607080900 10 20 30 40 50 60 70 80 90 100 110 120 130 140 150 160 170 180 190 200Depth (ft)Predicted Pile Capacity (tons)Project Number: Title: Squamish Shoppers Drug Mart Date: Parish: Station:      Offset: Elevation (ft): Pile Shape: End Bearing  & Ultimate CapacityRound CPT No: 2Pile Diameter: 1201020304050607080900 20 40 60 80 100 120 140 160 180 200 220 240Depth (ft)Predicted Pile Capacity (tons)Project Number: Title: Squamish Shoppers Drug Mart Date: Parish: Station:      Offset: Elevation (ft): Pile Shape: End Bearing  & Ultimate CapacityRound CPT No: 2Pile Diameter: 1601020304050607080900 20 40 60 80 100 120 140 160 180 200 220 240 260 280 300 320Depth (ft)Predicted Pile Capacity (tons)                   APPENDIX H: SITE PHOTOGRAPHS     Figure I: Front view of the pre-existing building.     Figure II: Pile driving hammer driving pile in place.     Figure III: Pile driving rig.      Figure IV: Front view of the constructed building near completion,        with compacted fill to grade in place.     Figure V: Front view of the completed building.     Figure VI: Side view of the completed building. 

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