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Minimum steel requirements for masonry walls : out-of-plane forces Gajer, Ruben Bernardo 1982

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CI MINIMUM STEEL REQUIREMENTS FOR MASONRY WALLS OUT-OF-PLANE' FORGES By RUBEN BERNARDO GAJER B . S c , Technion, I s r a e l I n s t i t u t e of Technology, 1977 THESIS SUBMITTED IN PARTIAL FULFILMENT OF THE REQUIREMENTS FOR THE DEGREE OF MASTER OF APPLIED SCIENCE i n THE DEPARTMENT OF CIVIL ENGINEERING We accept t h i s t h e s i s as conforming to the requi red standard THE UNIVERSITY OF BRITISH COLUMBIA January 1982 © Ruben Bernardo Gajer , 1982 In present ing t h i s t h e s i s i n p a r t i a l f u l f i l m e n t of the requirements f o r an advanced degree at the U n i v e r s i t y of B r i t i s h Columbia, I agree that the L i b r a r y s h a l l make i t f r e e l y a v a i l a b l e for reference and study. I fur ther agree that permission fo r extensive copying of t h i s t h e s i s f o r s c h o l a r l y purposes may be granted by the Head of my department or by h i s or her representa t i ves . It i s understood that copying or p u b l i c a t i o n of t h i s t h e s i s for f i n a n c i a l gain s h a l l not be allowed without my wr i t ten permis-s i o n . Department of C i v i l Engineer ing The Un ivers i ty of B r i t i s h Columbia 2324 Main Mal l Vancouver, B . C . , Canada V6T 1W5 R. B. Gajer - i i -ABSTRACT The behaviour of reinforced masonry walls subjected to out-of-plane forces, and the limitations on amounts and spacing of reinforcement required by the present code were examined. A research project was carried out in order to determine experiment-ally the capacity of the walls, the appropriate spacing of the main rein-forcing steel, and the effectiveness and appropriate spacing of transverse reinforcement. Full size non-loadbearing walls were tested under monotonic quasi-static loading. Test results showed that for typical 8 feet (2.44 m) storey height walls, the main steel may be spaced at more than 4 feet (1.22 m) and f u l l primary bending moment may s t i l l be achieved. Joint reinforcement appeared to be effective as distribution steel or as main steel for horizontal spans. - i i i -TABLE OF CONTENTS P a 9 e ABSTRACT i i TABLE OF CONTENTS i i i LIST OF TABLES v i LIST OF FIGURES v i i i ACKNOWLEDGEMENT i x CHAPTER 1. INTRODUCTION 1 1.1 Scope 2 2. STATE OF ART 5 2.1 Bending of Masonry Walls 5 2.2 Wal ls Under Combined Bending and Compression Loads 10 3. CURRENT ANALYSIS METHODS AND CODE APPROACH 25 3.1 Ana lys is of Walls Under Combined Bending and A x i a l Loads 25 3.1.1 P l a i n Masonry 25 3.1.2 E l a s t i c A n a l y s i s of Reinforced Masonry Walls 27 3.1.2(a) Uncracked Sect ion 27 3.1.2(b) Cracked Sect ion 28 3 . 1 . 2 ( b . l ) S tee l in Compression 28 3 .1 .2(b.2) S tee l i n Tension 28 3.1.3 Ult imate Strength Ana lys is 28 3.2 Code Design Procedures 29 3.2.1 Loadbearing Walls 29 3.2.1(a) The ACI (Ref .21) , NCMA (Ref.22) 29 and UBC (Ref .23, only f o r Reinforced Masonry) Codes. 3.2.1(b) Commentaries on Part 4 of the 1975- 32 NBC(Ref.24), CSA-CAN3-S304 (Ref .20) , 1969 SCPI: B u i l d i n g Code Requirements for Engineered Br ick Masonry (Ref .25) , UBC Code (Ref. 23, only f o r Unreinforced Masonry). CHAPTER 3.2.1(c) Minimum Thickness fo r Loadbearing Walls 3.2.1(d) Minimum Reinforcement for Loadbearing Walls 3.2.2 Non-Loadbearing Walls 3.2.2(a) Minimum Thickness fo r Non-Loadbearing Walls 3.2.2(b) Minimum Reinforcement f o r Non-Loadbearing Walls 4. MATERIAL PROPERTIES 4.1 Concrete Block Units 4.1.1 Compressive Strength of Masonry Blocks 4.1.1(a) Block Compressive Strength Tests at the U n i v e r s i t y of B r i t i s h Columbia 4.1.2 Block Tens i l e Tests 4.1.2(a) Block Tens i l e Tests at the U n i v e r s i t y . of B r i t i s h Columbia 4.2 Mortar 4.2.1 Mortar Tests at the U n i v e r s i t y of B r i t i s h Columbia 4.2.1(a) Compression tes t 4.2.1(b) Tensi le -Bond Strength Tests at the Un ive rs i t y of B r i t i s h Columbia 4.3 Grout 4.3.1 Grout Compression Tests at the U n i v e r s i t y of B r i t i s h Columbia 4.4 Determination of the Compressive Strength of Masonry ( f m ) 4.4.1 Prism Tests 4.4.1(a) Var iab les In f luencing Prism Compressive St rength . 4.4.2 Uni t Test Method 4 .4 .3 Prism Test at the U n i v e r s i t y o f B r i t i s h Columbia - v -P a ? e CHAPTER 4.5 Re in forc ing Stee l 58 4 . 5 . 1 . Re in forc ing Bars 58 4.5.1(a) Tests at the U n i v e r s i t y of B r i t i s h 59 Columbia. 4.5.2 J o i n t Reinforcement 59 4.5.2(a) Tests at the Un ive rs i t y of B r i t i s h 59 Columbia 5. TEST SERIES AT THE UNIVERSITY OF BRITISH COLUMBIA 67 5.1 Test Set-up 68 5.2 Test Ser ies 69 5.3 Test Results 69 5.3.1 V e r t i c a l Spans 69 5.3.2 Hor izonta l Spans 72 5.4 Ana lys is of Results 74 5.4.1 Bending 7 4 5.4.1(a) F i r s t Cracking 74 5.4.1(b) Capaci ty of P r i n c i p a l Reinforcement 75 5.4.1(c) Bending Resistance of Blocks Between 79 Reinforcement 5.5 Shear and Bond ( V e r t i c a l Spans) 85 6. SUMMARY AND CONCLUSIONS 111 REFERENCES 113 - v i -LIST OF TABLES TABLE Page 4.1 Compression Strength of Concrete Blocks 60 4.II Tension Tests of Blocks " 60 4 . I l l Part I - Mortar Compressive Strength 61 Part II- Mortar Compressive Strength 62 (Samples Corresponding to Wall Ser ies Not Tested i n This Program) 4.IV Part I - Mortar Tensi le -Bond Strength 63 Part II- Mortar Tensi le -Bond Strength 63 (Samples Corresponding to Wall Ser ies Note Tested i n Th is Program) 4.V Part I - Grout Compressive Strength 64 Part II- Grout Compressive Strength 64 (Samples Corresponding to Wall Ser ies Not Tested Yet i n the Program) 4.VI(a) Prism Tests 65 4.VI(b) Prism Tests Capped With Dona Cona Boards 65 4.VII Tension Tests of Re in forc ing Bars 66 4.VIII Tension Tests of Wire Reinforcement 66 5.1 D e t a i l s of Results of Tests on Masonry Wall Panels 88 5.II Y i e l d A n a l y s i s Results 9 2 5 . I l l Shear Stresses - Wall (1) 93 5.IV Bond Stresses - Wall (1) 93 - v i i -LIST OF FIGURES FIGURE Page 2.1 Loading Condi t ions and Moment D i s t r i b u t i o n Assumed i n 22 Ref. 9 2.2 In teract ion Curve fo r Short Pr ismat ic Wall 23 2.3 P-M In terac t ion Curve 24 3.1 End and Loading Condi t ions Assumed by (a) Y o k e l ' s Ser ies 4 1 of Test i n Ref . 10 and (b) 1969 SCPI Code. 5.1 Test Arrangement 94 5.2 Cent ra l Displacement v s . Load - Wall (14) 95 5.3 Cent ra l Displacement v s . Load - Wall (9) 95 5.4 Cent ra l Displacement v s . Load - Wall (10) 96 5.5 Cent ra l Displacement v s . Load - Wall (11) 96 5.6 Cent ra l Displacement v s . Load - Wall (12) 97 5.7 Cent ra l Displacement v s . Load - Wall (13) 9 7 5.8 Cent ra l Point Displacement i n a Hor izonta l and i n a 98 V e r t i c a l Plane (Wall 10) 5.9 Cent ra l Point Displacement i n a Hor izonta l and i n a 99 V e r t i c a l Plane (Wall 11) 5.10 Wall (11) Load v s . Cent ra l Displacement for one Complete 10 0 Cycle 5.11 Load-St ra in P lo t for Re in forc ing Stee l i n Wall (4) 101 5.12 Load-St ra in P l o t for Re in forc ing Stee l i n Wall (6) 10 2 5.13 Load-St ra in P l o t for Re in forc ing Stee l i n Wall (7) 10 3 5.14 Load-St ra in P lo t for Re in forc ing S tee l i n Wall (8) 104 5.15 Load-St ra in P lo t fo r Re in forc ing S tee l i n Wall (14) 105 5.16 Load-St ra in P lo t for Re in forc ing Stee l i n Wall (9) 106 5.17 Load-St ra in P lo t for Re in forc ing Stee l i n Wall (10) 107 5.18 Mechanism of F a i l u r e for Masonry Wall Spanning 108 H o r i z o n t a l l y - v i i i -FIGURE 5.19 Shear Stress D i s t r i b u t i o n on Shear Area 5.20 Sect ion i n V e r t i c a l Bending 5.21 Hor izonta l Bending Mechanism 5.22 E f f e c t i v e Shear Area for Out -o f -P lane Forces - i x -ACKNOWLEDGEMENT I would l i k e to express my grat i tude to Dr. D.L. Anderson, who served as my major a d v i s o r , fo r h is continued i n t e r e s t , u n f a i l i n g help and for h is many valuable suggestions to improve the presenta t ion of t h i s t h e s i s . I would a lso l i k e to thank Dr. N.D. Nathan f o r h i s advice and many h e l p f u l d iscuss ions at the ea r l y stages of my work. I am a lso indebted to Dr. S. Cherry f o r h i s continuous support and h is cons t ruc t ive comments on t h i s t h e s i s . I am very g r a t e f u l to the Masonry Research Foundation of Canada whose f i n a n c i a l a i d made my work p o s s i b l e , and to Mr. R . J . Barnes of Ocean Construct ion Suppl ies L imi ted and Mr. A.W. Weston f o r t h e i r i n t e r e s t i n my work. Vancouver, January, 1982 1. CHAPTER 1 INTRODUCTION Masonry i s amongst the o ldest but p o s s i b l y the l e a s t understood o f b u i l d i n g mater ia ls i n widespread use today. Th is lack of knowledge, mainly of i t s s t r u c t u r a l behaviour , has led to the misuse o f masonry through i n a -dequate or even non-ex is tent design procedures and poor cons t ruc t ion prac-t i c e . H i s t o r i c a l l y masonry s t ructures were unre in forced and gained t h e i r f l e x u r a l s t rength through g rav i ty l o a d s . During the 1930's the use of r e i n f o r c e d masonry was in t roduced, permi t t ing the cons t ruc t ion of more slender s t r u c t u r e s . Some research work s tar ted i n a number of countr ies i n Europe and North America l a te in the 1920's and ea r ly i n the 1930*s. Engineered masonry c o n s t r u c t i o n , based on engineer ing a n a l y s i s , was i n t r o -duced i n Codes dur ing the 1960's when the American Standard A s s o c i a t i o n (ASA) B u i l d i n g Code Requirements for Reinforced Masonry appeared, although r e i n f o r c e d masonry had been s p e c i f i e d s ince 1935 i n the Uniform B u i l d i n g Code (UBC). The new p r o v i s i o n s in, the sec t ion on masonry design i n the 1965 Nat ional B u i l d i n g Code (NBC) of Canada, on the s t r u c t u r a l a n a l y s i s of r e i n -forced and loadbear ing masonry, marked the beginning of engineered masonry i n t h i s country . The sec t ion on masonry design was rev ised for the 1970 e d i t i o n of NBC and again fo r i t s 1975 e d i t i o n . The new Canadian Standard A s s o c i a t i o n (CSA) Standard Can3-S304 Code was produced by a j o i n t e f f o r t o f the CSA and NBC. The i n t r o d u c t i o n o f r e i n f o r c e d masonry al lowed the use of th inner wal l s e c t i o n s , created the a b i l i t y to span over openings, and genera l l y made masonry a more economical b u i l d i n g m a t e r i a l . Today, f o r example, r e i n -forced masonry i s used as the s t r u c t u r a l support i n high r i s e b u i l d i n g s located i n s e i s m i c a l l y a c t i v e zones. Reinforced masonry i s thus a r e l a t i v e l y new s t r u c t u r a l mater ia l and only r e c e n t l y has there been any appreciable research i n t o i t s behaviour . Many c lauses of the codes present ly i n force have been produced by apply ing theor ies of r e i n f o r c e d concrete to r e i n f o r c e d masonry, and then in t roduc ing la rge f a c t o r s of sa fe ty to account fo r the unknown behaviour o f the masonry. Other c lauses i n the codes seem to have ne i ther an experimental nor an analogous b a s i s behind t h e i r requirements. Thus, there are many subjects i n r e i n f o r c e d masonry that requi re experimental research and the development of s u i t a b l e a n a l y t i c a l methods before the f u l l c a p a b i l i t y of the mater ia l can be s a f e l y u t i l i z e d . 1.1 Scope In t h i s t h e s i s the behaviour of a r e i n f o r c e d masonry wal l subjected to l a t e r a l loads ( loads app l ied perpendicular to the wal l surface) i s i n v e s t i -gated. L a t e r a l loads a r i s i n g from earthquake e x c i t a t i o n are the main con-s i d e r a t i o n , but the r e s u l t s could be app l icab le to l a t e r a l loads a r i s i n g from wind or mechanical means. When r e i n f o r c i n g a wal l to sus ta in l a t e r a l loads (ou t -o f -p lane bend-ing) , the d i r e c t i o n of the main bars w i l l depend upon the support c o n d i -t i o n s . The quant i ty o f main reinforcement can be e s t a b l i s h e d fo l lowing wel l known p r i n c i p l e s i f the transverse forces and the boundary cond i t ions are known. The design of the d i s t r i b u t i o n s t e e l and i t s in f luence on the maximum acceptable spacing of the p r i n c i p a l reinforcement are not as amenable to a n a l y s i s . Thus, we can say that the d e s i g n e r ' s problem i s not how much p r i n c i p a l s t e e l to use , but when and how to p lace the d i s t r i b u t i o n s t e e l i n order to get the maximum spacing of the p r i n c i p a l re inforcement. The minimum amount and maximum spacing of s t e e l i n r e i n f o r c e d masonry wal ls i s genera l ly s p e c i f i e d in codes without cons idera t ion of the l o c a t i o n i n the b u i l d i n g , or the conf igura t ion of the w a l l , and i n some cases the expected l e v e l of seismic a c t i v i t y ; they are based on s u c c e s s f u l past p r a c -t i c e . Those s t e e l requirements represent a considerable economic fac tor in the use of masonry and have given r i s e to controversy among the engineer and the c o n t r a c t o r . The aim of t h i s t h e s i s i s to examine the l i m i t a t i o n s on amounts and spacing of reinforcement required by the present Canadian Code> when the wal l i s loaded by o u t - o f - p l a n e f o r c e s . A review of some of the past research work c a r r i e d out on p l a i n and r e i n f o r c e d wa l ls tes ted under i n -plane compression, f l e x u r a l loading and t h e i r combinat ions, i s presented. The theor ies used i n the ana lys is and design of p l a i n and r e i n f o r c e d masonry, as wel l as the procedures f o r designing masonry wal ls under d i f f e r e n t codes of p r a c t i c e s are examined. The p roper t i es of i n d i v i d u a l mater ia ls and masonry assemblages, of importance i n the design of masonry as wel l as i n research work, are s tudied and the experimental data obtained on the mater ia ls used i n the U n i v e r s i t y of B r i t i s h Columbia s e r i e s of t e s t s are presented. The t e s t s on masonry wal ls descr ibed i n t h i s work are par t o f an experimental program design to determine the ac tua l minimum amounts and maximum spacing of s t e e l necessary to provide the strength required for the var ious seismic zones. The long term program w i l l take the fo l lowing form: (a) Study of r e i n f o r c e d masonry wal ls under monotonic q u a s i - s t a t i c loading to f a i l u r e to determine the strength and load-deformat ion character -i s t i c s of the wal l panels with var ious amounts and arrangements of the s t e e l re inforcement . (b) The r e s u l t s obtained i n (a) to be confirmed by q u a s i - s t a t i c c y c l i c load ing . (c) Design procedures emerging from (a) and (b) to be v e r i f i e d by t e s t i n g under simulated earthquake load ing on a shaking t a b l e . To date the t e s t s have involved only monotonic q u a s i - s t a t i c l a t e r a l loads with the except ion of 2 wal ls which were rotated and subjected to one q u a s i - s t a t i c load r e v e r s a l . The immediate aims o f the present study a re : 1) To e s t a b l i s h the maximum spacing of main s t e e l without any t ransverse d i s t r i b u t i o n s t e e l , f o r wal ls spanning e i ther h o r i z o n t a l l y or v e r t i -c a l l y . 2) To determine the e f f e c t of the d i s t r i b u t i o n s t e e l on the spacing of the main s t e e l . 3) To determine the e f f i c i e n c y of j o i n t reinforcement as h o r i z o n t a l d i s t r i b u t i o n s t e e l or as main s t e e l for h o r i z o n t a l l y spanning w a l l s . 4) To determine a method of p r e d i c t i n g the a b i l i t y of the masonry to span between the main reinforcement or l a t e r a l supports . 5. CHAPTER 2 STATE-OF-THE-ART This chapter presents a review of some of the research work which has helped to form the b a s i s of the current a n a l y s i s and design of masonry w a l l s . Concrete masonry i s p r i m a r i l y a load-bear ing mater ia l with a r e l a t i v e -l y high compressive s t rength . The t e n s i l e strength i s low and h igh ly v a r i a b l e . Massive masonry wal ls have the a b i l i t y to r e s i s t l a t e r a l loads by g rav i ty s t a b i l i t y , but very l i t t l e information has been a v a i l a b l e to enable engineers to design t h i n wal ls f o r l a t e r a l l o a d s . Since masonry has a very low t e n s i l e strength and i s b r i £ t l e i n t e n -s i o n , reinforcement i s needed i n tens ion areas . Reinforced masonry wal ls are reasonably d u c t i l e and t h e i r behaviour i s qu i te s i m i l a r to that of r e i n f o r c e d concre te . Thus, r e i n f o r c e d masonry design i s based upon theor-i e s long used i n des igning r e i n f o r c e d concrete elements. Studies are being done to incorporate the more soph is t i ca ted concepts of u l t imate strength and l i m i t s ta te d e s i g n . Work i n the laboratory i s being helped by the use of the F i n i t e Element a n a l y s i s to reproduce, i n t e r p r e t or p r e d i c t behaviour of wal ls dur ing t e s t s . 2.1 Bending of Masonry Walls The strength of masonry wal ls subjected to l a t e r a l loads depends upon the h o r i z o n t a l and v e r t i c a l f l e x u r a l s t rength , which i s a f f e c t e d by i n -plane forces e i t h e r from loads or arching ac t ion due to edge r e s t r a i n t , the amount and arrangement of r e i n f o r c i n g s t e e l , and the st rength of the 6. masonry components. Shear and bond f a i l u r e s may a lso cause o v e r a l l f a i l -u re . A repor t on s i x s i n g l e wythe concrete b lock wal l panels tes ted i n pure h o r i z o n t a l f lexure was presented by Cox and Ennenga (Ref. 1 ) . They be l i eved that the b u i l d i n g code requirements of t ransverse support provided at e i ther h o r i z o n t a l or v e r t i c a l i n t e r v a l s not exceeding 18 times the nominal wal l t h i c k n e s s , though s a t i s f a c t o r y f o r wa l ls without openings, could not be app l ied regard less of the wal l shape and, i n the case of wal ls with openings, the s i ze and arrangement of those openings. Three panels were b u i l t of 8 x 8 x 16 i n . (200 x 200 x 400 mm) hollow concrete b l o c k s . Three a d d i t i o n a l w a l l s , b u i l t i n the same f a s h i o n , had standard j o i n t reinforcement (#9-gauge wi re ) . A l l the specimens were 8 f t . (2.44 m) long . The r e i n f o r c e d wal ls were 5 courses h i g h , the unre inforced ones were 3, 5 and 7 courses h igh . A l l the panels were tested under c o n d i -t ions of simple suppor ts , spanning h o r i z o n t a l l y with a concentrated l i n e load i n the center . Test r e s u l t s i n d i c a t e d that the wal ls cracked at bending moments rang-ing from 1140 f t l b / f t (256 Nm/m) to 1280 f t l b / f t (287 Nm/m). The modulus of rupture , based on the gross area was about 110 p s i (0.76 MPa). The reinforcement d id not in f luence the load under which the wal l c racked, but i t d id act to c o n t r o l the cracks and to preserve the wal l a f te r i t c racked. In 1961 Hedstrom (Ref. 2) reported a s e r i e s of t e s t s on concrete masonry w a l l s . The program was aimed at determining: (a) the f l e x u r a l strength of v e r t i c a l spanning w a l l s , (b) the e f f e c t of h o r i z o n t a l j o i n t reinforcement on the st rength o f a h o r i z o n t a l l y spanning wal l and (c) the e f f e c t of v e r t i c a l loads on the f l e x u r a l st rength of wal ls spanning v e r t i c a l l y . The running bond specimens were 4 x 8 f t . (1.22 x 2.44 m) and 8 x 4 f t . (2.44 x 1.22 m), b u i l t with 8 x 8 x 16 i n (200 x 200 x 400 mm) u n i t s . Some of the wal ls spanning h o r i z o n t a l l y were r e i n f o r c e d with #9 gauge j o i n t reinforcement i n the h o r i z o n t a l mortar j o i n t s (some at 8 i n . on cente r , others at 16). From h i s t e s t s r e s u l t s the author concluded tha t : (a) for v e r t i c a l spans, wal ls b u i l t with type M or S mortars showed mortar bond strengths of 60 p s i (0.41 MPa) and 30 p s i (0.20 MPa) r e s p e c t i v e l y , (b) for hor i zon ta l spans, the p a r t i c u l a r j o i n t reinforcement used d i d not s i g n i f i c a n t l y increase the load at f i r s t c rack ing but i t was e f f e c t i v e i n i n c r e a s i n g the ul t imate strength of the w a l l s ; the r a t i o of c rack ing load i n the h o r i z o n -t a l d i r e c t i o n to the one i n the v e r t i c a l d i r e c t i o n ranged from 2 to 4, (c) the add i t ion of a v e r t i c a l compressive load to the wal ls spanning v e r -t i c a l l y s i g n i f i c a n t l y increased t h e i r f l e x u r a l s t rength . Fishburn (Ref. 3) performed f l e x u r a l t es ts on 34 concrete masonry w a l l s . Twenty-eight of the wal ls had normal h o r i z o n t a l bed j o i n t s , but 6 wal ls were constructed so that the bed j o i n t s were v e r t i c a l . The supports cons is ted of simply supported top and bottom r e a c t i o n s , and the l a t e r a l load was app l ied through l i n e loads at the quarter p o i n t s . The wal l specimens were 4 fee t long (1.22 m) by 8.67 f t . high (2.64 m) b u i l t with 8 x 8 x 16 i n . (200 x 200 x 400 mm) concrete b l o c k s . The modulus of rupture of the wal ls tes ted with the bed j o i n t s normal to the span, based on the gross area as given by the authors ranged from 10 p s i (0.09 MPa) to 24 p s i (0.17 MPa) i n wal ls with type N mortar and from 17 • p s i (0.12 MPa) to 33 p s i (0.23 MPa) with type S mortar . The f l e x u r a l st rength of the wal ls tes ted with the bed j o i n t s p a r a l l e l with the span was about 3 times h igher . 8. Sah l in (Ref. 4) reported an i n v e s t i g a t i o n by N i l s s o n (Ref. H.23 i n Ref . 4) where the modulus of rupture i n h o r i z o n t a l bending was found to be 3 to 6 times higher than the modulus of rupture i n v e r t i c a l bending. The importance of r e s t r a i n i n g the v e r t i c a l edges of plane masonry wal ls against r o t a t i o n and in -p lane l a t e r a l displacement was studied by C. Anderson (Ref. 5) and by C. Anderson and N . J . Br ight (Ref. 6 ) . As reported i n Ref. 5 s ix concrete block wal ls were t e s t e d . The base of a l l the wal ls were b u i l t on a mortar bed to provide p a r t i a l r o t a t i o n a l r e s t r a i n t , and the tops were f r e e . Three of the wal ls were simply supported along the v e r t i -c a l edges while the other three were b u i l t i n t o 3.3 foo t (1.0 m) long crosswal ls which in turn were t i e d to the reac t ion f l o o r . A second s e r i e s of t e s t s (Ref. 6) cons is ted of 5 concrete wal ls and one composite w a l l , a l l b u i l t with the same type of blocks used in the previous s e r i e s and con-s t ruc ted wi th in a s t e e l framework that could provide both r o t a t i o n a l and in -p lane edge r e s t r a i n t . The authors (Ref. 6) observed that when the ver -t i c a l supports provided moment r e s t r a i n t and an arching type r e a c t i o n , the wal ls could support much higher l a t e r a l l o a d s . From t h i s the authors con-cluded that any theory that attempts to p r e d i c t the strength of p r a c t i c a l wa l ls must be -capable of i n c l u d i n g the e f f e c t of edge r e s t r a i n t . A theory based simply on the bending capaci ty of unrest ra ined wal ls i s not adequate. West et a l . (Ref. 7) descr ibed f l e x u r a l t e s t s on more than 1000 small masonry samples (wal let tes) and on more than 10 0 f u l l - s i z e d wal ls up to 18 f t . (5.5 m) long and 11.8 f t . (3.6 m) h i g h , under uniform l a t e r a l l o a d . In t h e i r r e p o r t , the wr i te rs def ined the orthogonal r a t i o as the r a t i o of the f l e x u r a l strength for h o r i z o n t a l spans to that for v e r t i c a l spans. The small wal l specimens, tes ted under four po in t loading were used to determine the f l e x u r a l strength i n the v e r t i c a l and h o r i z o n t a l spans. In t h e i r t e s t s the authors obtained orthogonal r a t i o s of about 3. B .A. Hase l t ine et a l . (Part II, Ref. 7) compared the experimental r e s u l t s with some t h e o r e t i c a l methods of a n a l y s i s , assuming that the wal ls behave as p l a t e s . One of the methods used was y i e l d l i n e a n a l y s i s . The authors pointed out that one of the major advantages of t h i s theory i s that i t i s p o s s i b l e to e a s i l y feed in to the c a l c u l a t i o n s d i f f e r e n t strengths i n two orthogonal d i r e c t i o n s , enabl ing the bending r e s i s t a n c e of wal ls to be worked out us ing the ac tua l r a t i o of s t rengths . It i s a l s o p o s s i b l e to take in to account any bending res is tance over a support . L a t e r a l pressures p red ic ted using y i e l d l i n e theory were compared to the experimental f a i l u r e pressures . The moment c a p a c i t i e s were assumed to be p ropor t iona l to the strengths determined by the wa l le t te t e s t s , and edge r e s t r a i n t s as measured experimental ly were inc luded i n the c a l c u l a t i o n s . A p l o t of the experimental pressures v s . the c a l c u l a t e d pressures showed scat tered r e s u l t s , the p r e d i c t i o n being good for long wal ls but unconserva-t i v e f o r short w a l l s . A design method, based on y i e l d l i n e a n a l y s i s , was recommended by the authors. They suggested that the p a r t i a l end r e s t r a i n t be replaced with e i ther f u l l f i x i t y or no r e s t r a i n t , depending on whether the wal ls were continuous or simply supported. C h a r a c t e r i s t i c f l e x u r a l strength of b r i c k -work was given as a funct ion of the water absorpt ion of the b r i c k s and the type of mortar , and an orthogonal r a t i o of 3 was recommended. Ca lcu la ted pressures were p l o t t e d against the ac tua l f a i l u r e p r e s s u r e s . A l l the r e s u l t s were conserva t i ve , which was explained by the authors by p o i n t i n g out that i n t h i s case c o n t i n u i t y ( i . e . , edge r e s t r a i n t ) of the wal ls was ignored and f l e x u r a l strengths were smoothed out by us ing a c h a r a c t e r i s t i c v a l u e . 10. Scr ivener (Ref. 8) reported on two s e r i e s of t es ts of r e i n f o r c e d b r i c k wa l ls subjected to uniform l a t e r a l l o a d . In the f i r s t s e r i e s the wal ls were simply supported, but they were rotated to l i e i n the h o r i z o n t a l plane and so the wal l dead load was i n c o r r e c t l y a p p l i e d . A second s e r i e s was then run i n which the wal ls were kept in the v e r t i c a l p o s i t i o n and spanned v e r t i c a l l y with simple supports . A few load r e v e r s a l s , by apply ing the load to the other f a c e , were a lso c a r r i e d out . Scr ivener found that i n both s e r i e s , assuming a l i g h t l y r e i n f o r c e d wide beam s e c t i o n ( i e . the whole wal l c r o s s - s e c t i o n as the beam sect ion and assuming no f a i l u r e between bars) and apply ing ul t imate strength theory as fo r r e i n f o r c e d concre te , the tes t y i e l d load could be p red ic ted wi th in a smal l d e v i a t i o n . The wa l ls presented a h igh ly d u c t i l e behaviour character -i zed by large i n e l a s t i c deformations. 2.2 Walls Under Combined Bending and Compression Loads Many wal ls are subjected to both v e r t i c a l a x i a l loads and bending moments which are caused by e i ther l a t e r a l loads or eccen t r i c a x i a l l o a d s . In h i s repor t on t e s t s on 4 f t . x 8 f t . x 8 i n . (1.22 m x 2.44 m x 200 mm) concrete masonry w a l l s , Hedstrom (Ref. 2) found that with few excep-t ions wal ls subjected to an eccen t r i c v e r t i c a l load with e = t / 6 , where e i s the load e c c e n t r i c i t y and t the th ickness of the w a l l , f a i l e d at 35 to 50% of the strength of the i n d i v i d u a l b l o c k s . He a lso observed that the compressive strength of the wal ls depended p r i m a r i l y on the strength of the b locks and was l i t t l e a f fec ted by the strength of the mortar . For wal ls subjected to a v e r t i c a l p lus l a t e r a l l o a d , he found that a f t e r the f i r s t c rack ing the precompressed wal ls continued to ca r ry i n c r e a s i n g l a t e r a l loads up to about 1.25 times the c rack ing l o a d . 11. Fishburn (Ref. 3) performed a s e r i e s of 26 tes ts on 4 f t x 8 f t x 8 i n (1.22 m x 2.44 m x 200 mm) wal ls loaded with a v e r t i c a l load app l ied with an e c c e n t r i c i t y of t / 6 . The author concluded that the compressive st rength of the concrete masonry wal ls increased only s l i g h t l y with i n c r e a s i n g mortar compressive s t rength . Based on the net a r e a , the strength o f the wa l ls was about h a l f of the block s t rength . In 1970 Yokel et a l . (Ref. 9) reported on a research program aimed at determining the e f f e c t s of slenderness and load e c c e n t r i c i t y on the strength of s lender concrete masonry w a l l s . The specimens inc luded 32, 8 i n . (200 mm) unre in forced concrete masonry wal ls and 28, 6 i n . (150 mm) r e i n f o r c e d concrete masonry w a l l s . The specimens were 4 f t . (1.22 m) wide by 10 f t . (3.05 m), 16 f t . (4.88 m) and 20 f t . (6.10 m) h i g h . The wal ls were tes ted to des t ruc t ion under compressive loads app l ied a x i a l l y and at e c c e n t r i c i t i e s of 1/6, 1/4 and 1/3 of the wal l t h i c k n e s s . The tes t set up was designed to prevent r o t a t i o n at the bottom while a l lowing r o t a t i o n at the top. Three and two-block high prisms b u i l t i n stacked bond were t e s t e d , under the same loading condi t ions that were used for the f u l l scale s p e c i -mens. The pr ism t e s t s ind ica ted that f l e x u r a l compressive strength increased with i n c r e a s i n g s t r a i n gradients ( i . e . e c c e n t r i c i t y ) . The f l e x -u r a l compressive s t rength was def ined as a f ' , where f ' i s the u n i a x i a l m m compressive strength of the masonry, and a i s a func t ion of e ( i . e . the s t r a i n gradient) and a > 1.0. There was a good c o r r e l a t i o n between short wa l l (10 f t ) strength and prism strength under concent r ic l o a d . Under e c c e n t r i c load the prisms 12. developed higher compressive strength than the r e i n f o r c e d wal l specimens which f a i l e d near the t o p , where the e c c e n t r i c load was a p p l i e d . The authors concluded that f a i l u r e in t h i s region i s an i n d i c a t i o n that s l e n -derness had no s i g n i f i c a n t e f f e c t on these short w a l l s . The discrepancy between the f l e x u r a l strength of the prisms and wal ls i s a t t r i b u t e d to poor composite ac t ion amongst the grout , b locks and reinforcement of the w a l l s . The authors recommended the in t roduct ion of the moment magnif ier method in c a l c u l a t i n g the moments ac t ing on slender masonry w a l l s . The maximum moment should be approximated by: M = PeC / ( 1 - P / P ) > Pe, where m cr C = 0 . 6 + 0 . 4 M./M_, a f a c t o r dependent upon d i f f e r e n t end condi t ions (M m l 2 - . 1 and M 2 are the end moments and |M^|>| M 2 | ) , P i s the a x i a l l o a d , e i s the l a r g e s t e c c e n t r i c i t y , and P = n 2 E I / ( k h ) 2 . The magnitude of t h i s moment magni f ier e f f e c t i n the case of masonry depends on severa l parameters: 1) end f i x i t y : f l a t ended condi t ions at the base of the wal ls gave on a minor amount of f i x i t y . For such a cond i t ion the authors assumed end moments and an e f f e c t i v e length as shown i n Figure 2 .1 . 2) S t i f f n e s s E I : E decreases with i n c r e a s i n g s t resses and I decreases i f the s e c t i o n c r a c k s . For the r e i n f o r c e d wal ls the equivalent EI was approximated by EI = E.I / 2 . 5 , where l n E^ i s the i n i t i a l tangent modulus of e l a s t i c i t y and 1^ i s the moment of i n e r t i a of the uncracked s e c t i o n . For the unre inforced wal ls EI was taken as E I / 3 . 5 . l n For the 6 i n . (150 mm) wal ls a short wal l i n t e r a c t i o n curve fo r the sec t ion capaci ty was developed on the b a s i s of the average a x i a l strength of the prisms ( f * ) . From t h i s curve , i n t e r a c t i o n curves fo r slender wal ls m were developed by reducing the moment at each l e v e l of P by the moment magnif ier f a c t o r . I t was concluded tha t , except for the case of the 20 f t . 13. (6.10 m) wal ls with e = t / 3 , the t h e o r e t i c a l i n t e r a c t i o n curves were conserva t ive . A new wal l i n t e r a c t i o n curve was c a l c u l a t e d on the b a s i s of the f l e x u r a l strength of the prisms for e = t / 3 . In t h i s case the trend of the t e s t s r e s u l t s and the ac tua l f a i l u r e loads were i n good agreement with the t h e o r e t i c a l p r e d i c t i o n s . In the case of the 8 i n . (200 mm) wal ls the conc lus ion was that the strength of s lender wal ls was. conserva t ive ly p red ic ted by the moment magni-f i e r method when basing the c a l c u l a t i o n s on the average a x i a l pr ism s t rength . The magni f ica t ion of the moments, as wel l as the strength of s lender wa l ls were approximately p red ic ted by the moment magnif ier method, when the f l e x u r a l compressive strength at load e c c e n t r i c i t i e s greater than t /3 was assumed to equal the average f l e x u r a l strength of the prisms loaded at e = t / 3 . In 1971, Yokel et a l . (Ref. 10) publ ished a repor t on a second ser ies of tes ts on 90 wal ls of 10 d i f f e r e n t types of masonry cons t ruc t ion using both b r i c k s and concrete b locks under var ious combinations of v e r t i c a l and t ransverse l o a d . The purpose of t h e i r program-was to develop a n a l y t i c a l procedures to p r e d i c t the strength of masonry wal ls subjected to combined a x i a l and l a t e r a l loads . Both the a x i a l and l a t e r a l loads were appl ied un i formly . The concrete wa l ls were 4 x 8 f t . (1.22 x 2.44 m) b u i l t i n running bond wi th ; 8 x 8 x 16 i n . (200 x 200 x 400 mm) blocks with e i ther type N or high-bond mortars . In order to analyze t h e i r t e s t r e s u l t s the authors der ived the equa-t ions requi red to const ruct an i n t e r a c t i o n curve . The f l e x u r a l t e n s i l e 14. strength of masonry was assumed to be low compared with the f l e x u r a l compressive s t rength . The l a t t e r was assumed, as i n Ref . 9, to be equal to f ' . A l i n e a r s t ress approximation to the s t ress block was assumed. F igure m 2.2 shows t h e . i n t e r a c t i o n curve for a short p r ismat ic w a l l . The authors a l s o der ived the i n t e r a c t i o n equations for symmetrical and asymmetrical hollow s e c t i o n s . The moment magni f ier method, was used by the authors to account fo r the slenderness e f f e c t . The equivalent s t i f f n e s s was assumed as EI = E I i n (0.2 + P/P ) < 0.7 E. I and a lso as EI = E.I /3 (the l a t t e r g i v i n g more o l n l n conservat ive r e s u l t s ) . The top of the wal ls was assumed to be pinned while the bottom rested o n . f i b r e b o a r d . With these end condi t ions i t was assumed that the e f f e c t i v e length of the wal l (k£) was 0.8h and that the r e s u l t i n g moment at the base was equal to the maximum moment along the span and equal to 68 percent of the moment that would be obtained in the case of a pinned connection at the bottom of the w a l l . In order to make a meaningful comparison between the i n t e r a c t i o n curve (predicted for short wal ls) and the strength of a slender w a l l , the moment a t t r i b u t a b l e to d e f l e c t i o n s , approximated by the a x i a l load times the measured maximum d e f l e c t i o n , was added to primary moments. Reduced i n t e r -ac t ion curves were drawn using the moment magnif ier method. See Figure 2 .3 . Yokel e t a l . , concluded that the trend of the r e l a t i o n s h i p between v e r t i c a l loads and moments was c o r r e c t l y p r e d i c t e d , and that the order o f magnitude of the observed added moment due to slenderness e f f e c t s showed f a i r l y good agreement with the p red ic ted values c a l c u l a t e d using the moment magni f ier method. 15. In 1976 F a t t a l and Cattaneo (Ref. 11) reported a program aimed at cont inu ing the work s ta r ted by Yokel et a l . , (Refs . 9 and 10). Tests were c a r r i e d out on prisms and w a l l s , b u i l t of 4 i n . (100 mm) b r i c k s , 6 i n . (150 mm) concrete b locks and 10 i n . (250 mm) composite w a l l s . The tes ts inc luded 95 prisms and 56 w a l l s . The prisms were subjected to v e r t i c a l compressive loads at var ious equal top and bottom e c c e n t r i c i t i e s . The wal ls were tested under var ious combinations of t ransverse and v e r t i c a l l o a d . The t e s t set -up was the same as descr ibed by Yokel et a l , (Refs . 9 and 10). A l l the wal ls were constructed i n running bond to a height of approximately 8 f t . (2.44 m) and were nominally 32 i n . (0.81 m) wide. The observed s t r e s s - s t r a i n r e l a t i o n s h i p s were f a i r l y l i n e a r with a maximum s t r a i n at f a i l u r e ranging between 0.001 and 0.002. The authors showed that the empi r ica l r e l a t i o n s h i p for t h e , f l e x u r a l r i g i d i t y proposed i n Ref . 10 (EI =.E.I (0.2 + P/P ) < 0.7 E. I ) produced x n o i n an i n t e r a c t i o n diagram that gave pred ic ted values which were cons is ten t with experimental r e s u l t s for the 3 types of masonry wal ls considered i n t h e i r program. As i n previous works (Ref. 9 and 10) the authors found that the com-press ive strength in f l exure ( a f ) exceeds the compressive strength ( f ' ) m m developed i n a x i a l compression. Within the range of e c c e n t r i c i t i e s used (t/12 < e < t /3 ) the average value of the c o e f f i c i e n t "a" fo r concrete blocks was found to be 1.27 with l i t t l e v a r i a t i o n . Thus the hypothesis advanced i n the previous s tud ies (Refs. 9 and 10) that the compressive strength increases with f l e x u r a l s t r a i n gradient was, according to the authors , not conf i rmed. The quest ion remains of what happens i n the range 0 < e < t / 1 2 . 16. The authors concluded that in the case of w a l l s , where f lexure was induced by l a t e r a l loads s i n g l y or in combination with e c c e n t r i c a l l y appl ied compressive l o a d s , the agreement between theory (using the moment magni f ier method) and experiments cons t i tu ted a g e n e r a l i z a t i o n of the b a s i c theory proposed in Refs . 9 and 10. In 1980, R.H. Brown and F .N . Wattar (Ref. 12) presented the r e s u l t s of an a n a l y t i c a l program that reviewed the r e s u l t s of the 252 tes ts reported by Refs . 9, 10 and 11. The authors developed f a i l u r e r e l a t i o n s h i p s between app l ied a x i a l load and bending moment assuming: a) The s t r e s s - s t r a i n curve i s l i n e a r b) F a i l u r e takes p lace at a compressive s t r a i n of 0.002 c) The r a t i o f ' / f i s equal to 0.02 ( f = f l e x u r a l t e n s i l e t ra t s t rength of masonry) The equations der ived i n Ref. 10 were used to develop the i n t e r a c t i o n curves for s o l i d pr ismat ic rectangular s e c t i o n s . A numerical ana lys is method was used to def ine the i n t e r a c t i o n curve fo r hollow pr ismat ic r e c -tangular s e c t i o n s . A minimum e c c e n t r i c i t y of t /12 was used to account for imper fec t ions . The maximum or magnif ied moment at f a i l u r e of the 252 t e s t s were superimposed with the short wa l l i n t e r a c t i o n curve . The magnif ied moments were obtained by m u l t i p l y i n g the appl ied moments by the moment magnif ier f a c t o r : C / ( 1 - P / P ). The safety fac tor was def ined by the r a t i o between m c r the magnif ied moment and the moment capac i ty given by the short wal l i n t e r a c t i o n diagram at the same a x i a l load l e v e l . A s t a t i s t i c a l study was performed and a f t e r a l o g t ransformat ion the 17. fac tor of safety f i t t e d a near ly normal d i s t r i b u t i o n . Only 4 wal ls ( a l l hollow concrete b lock wal ls) would have been overestimated by the moment magnif ier method. The confidence with which t h i s method can be app l i ed without any a d d i t i o n a l load f a c t o r s , i s shown by the 99.99% p r o b a b i l i t y (almost ce r ta in ty ) of obta in ing a safety fac tor exceeding 1.0 (some wal ls have a safe ty f a c t o r greater than 10.0) . The authors a lso concluded that the fac tor of safety was independent of the e c c e n t r i c i t y . Ha tz in iko las et a l . , (Refs . 13, 14, 15, 16 and 17) have c a r r i e d out an extensive program studying the behaviour of e c c e n t r i c a l l y loaded w a l l s . They a l s o used the moment magnif ier method as a means to account f o r the slenderness e f f e c t . In order to obta in a value of P requi red fo r the eva luat ion of the cr magni f ica t ion f a c t o r , they approximated the s o l u t i o n given by Yokal (Ref. 18) for wal ls without t e n s i l e strength and i n s i n g l e curvature by: Per = 8II2 (0.5 - e / t ) 3 EI / h 2 (I = uncracked moment of i n e r t i a ) (Refs. 13, o o 15 and 16). If some t e n s i l e strength (f^) was to be cons idered: Per = 8H 2 (0.5 - e / t + £ / 2 t ) 3 EI / h 2 where K i s the distance from the po int of zero o s t r e s s to the end of the c rack . For the case in which the wal l i s subjec-ted to unequal end e c c e n t r i c i t i e s producing s i n g l e curvature , the wr i te rs recommend the use of the average e va lue . A method fo r determining the buck l ing load fo r the case of double curvature was a lso developed (Refs. 13 and 15). The authors pointed out that in t h i s case buck l ing w i l l a lso tend to occur i n the primary s ing le l o o p . Th is behaviour was substant ia ted by t e s t r e s u l t s . The c r i t i c a l buck l ing load would be: P = aEI / h 2 where X i s a buck l ing c o e f f i c i e n t c r o depending on the end e c c e n t r i c i t i e s (Ref. 15) . 18. The buck l ing load for r e i n f o r c e d masonry wal ls i n s i n g l e curvature with e / t < 1/3 can be evaluated as for p l a i n masonry. For la rger values of e / t the behaviour w i l l be a funct ion of the extent of crack ing and the transformed moment of i n e r t i a . A lower l i m i t of EI (1/2 - e/ t ) > EI /10 i s o o recommended for the f l e x u r a l s t i f f n e s s . The a n a l y s i s of r e i n f o r c e d wal ls i n double curvature can be c a r r i e d out s i m i l a r l y to the a n a l y s i s of p l a i n masonry w a l l s . The authors found that the experimental r e s u l t s were in f a i r l y good agreement with the a n a l y t i c a l va lues . In 1980, Oj inaga and Turkstra (Ref. 19) reviewed some of the l i m i t a -t ions of the moment magnif ier method. A second approach, commonly termed the P-A method, i n which moments due to a x i a l fo rces a c t i n g through d e f l e c -t ions are combined with elementary bending moments and then compared to the sec t ion c a p a c i t y , was inves t iga ted for unre inforced b r i c k and concrete block masonry. As i n the previous works a l i n e a r s t r a i n - s t r e s s r e l a t i o n s h i p was adop-ted . For hollow concrete b locks the i n i t i a l tangent modulus of e l a s t i c i t y was assumed to be: E = 440 f + 73600 p s i < 3000000 p s i based on the m r e s u l t s of 52 t e s t s . In te rac t ion curves fo r short columns were c a l c u l a t e d using the equa-t ions developed by Yokel et a l , (Ref. 10). A maximum f l e x u r a l compressive strength equal to f* was assumed. m The moment magni f ier theory was f i r s t s t u d i e d . The magni f ica t ion fac tor was def ined as before (Refs. 9, 10, 11) and i n d e f i n i n g Per = I I 2 EI/m(kh) 2 , a r i g i d i t y reduct ion f a c t o r m equal to 2.5 was used. Th is theory was compared with experimental data and found to have some 19. l i m i t a t i o n s , the major one being a tendency to be unconservative for s ing le curvature bending under r e l a t i v e l y large e c c e n t r i c i t i e s . The authors con-cluded that crack ing across the sect ion and along the height of the wal l could not be modelled by simply ad just ing the r i g i d i t y reduct ion f a c t o r , m. When compared to a t h e o r e t i c a l "exact" ana lys is assuming l i n e a r s t r e s s -s t r a i n , cracked regions and a moment curvature approach descr ibed by Sahl in (Ref. 4) the moment magnif ier method appeared to be conservat ive for small e c c e n t r i c i t i e s but become unconservative for la rge e c c e n t r i c i t i e s . Although the use of a r i g i d i t y reduct ion fac tor vary ing with load l e v e l could reduce these problems, t h i s s o l u t i o n could lead to p r a c t i c a l problems in des ign . The authors int roduced t h e i r beam-column a n a l y s i s i n which l a t e r a l d e f l e c t i o n s due to end moments, t ransverse l o a d s , and P-A e f f e c t s are com-puted. The maximum moment along the span i s found tak ing a l l those c o n t r i -but ions in to account. A x i a l load and maximum bending moment are then compared with the short wal l s t rength . Successfu l a p p l i c a t i o n of such an ana lys is requi res an estimate of the e f f e c t i v e r i g i d i t y EI . . . The i n i t i a l e f f tangent modulus can be estimated by best f i t equations as the one presented above. I must consider the uncracked sec t ion as wel l as the end i n e r -e f f t i a s f o r e i t h e r cracked or uncracked c o n d i t i o n s . The authors recommended a general approximation as fo l lows: (a) I - (I + I )/4 fo r 0 < e / e < + 1 e f f endl end2 1 2 (I + I ) /4 („> I e f f = minimum { ^ + I ) / 4 " fo r -1 < < 0 end2 where I , 1 , and I are the two end i n e r t i a s (which could be of endl end2 20. cracked sect ions) and the uncracked sec t ion i n e r t i a , r e s p e c t i v e l y and e^, e 2 are the end e c c e n t r i c i t i e s . The e f f e c t i v e length concept i s replaced by cons idera t ion of app l ied end moments and estimated de f l ec ted shape. In t h e i r a n a l y s i s , the wr i te rs considered only the e l a s t i c d e f l e c t i o n s due to primary bending moments, which caused underestimation of d e f l e c t i o n s but avoided i t e r a t i v e s o l u t i o n s . Th is type of a n a l y s i s w i l l not p r e d i c t buck-l i n g loads nor does i t g ive a good est imat ion of the maximum moments for small e c c e n t r i c i t i e s . , thus a l i m i t i n g slenderness r a t i o and minimum eccen-t r i c i t i e s were imposed. In order to evaluate the approximation i m p l i c i t i n the approach and e s t a b l i s h reasonable l i m i t s of a p p l i c a b i l i t y , the r a t i o of f a i l u r e load to a x i a l capac i ty v s . the slenderness r a t i o fo r d i f f e r e n t e c c e n t r i c i t i e s was p l o t t e d against the values obtained using the Sah l in method (Ref. 4 ) . As a r e s u l t of t h i s comparison a minimum e c c e n t r i c i t y of t /12 and a maximum slenderness r a t i o of 80 are recommended ( for a l l eccen-t r i c i t i e s > t / 1 2 ) . The authors recommended, i n the case of no l a t e r a l loads and short w a l l s , the use of the short wal l c a p a c i t y , without c a l c u l a t i o n of l a t e r a l d e f l e c t i o n s and slenderness e f f e c t s , w i th in the fo l low ing l i m i t s : (a) L / r = 35 - 17.5 ^ ^ ^ 2 fo r 0.0 < e 1 / e 2 < 1 , 0 (b) L / r = 35 - 35 e ^ A ^ fo r -1 .0 < e1/e2 < 0.0 where L i s the wal l height and r the radius of g y r a t i o n . Comparison was a lso made with experimental da ta . In t h i s case a m i n i -mum e c c e n t r i c i t y of O.Ol t was i n c l u d e d , ra ther than the minimum of t /12 recommended fo r des ign , because of the we l l c o n t r o l l e d c o n d i t i o n s . On average, the ana lys is was found to be conserva t ive . 21. A comparison i s made a lso with the procedures o u t l i n e d i n the present Canadian Code (Ref. 20). I t was found that the present code permits r e l a -t i v e l y higher t h e o r e t i c a l strengths at low s lenderness. In general the proposed approach may be more or l e s s conservat ive than the code depending on the slenderness r a t i o , e , / e ^ r a t i o and maximum e c c e n t r i c i t i e s . Loading Condi t ions and Moment D i s t r i b u t i o n Assumed i n Ref. 9. F igure 2.1 P / P o t 1.0 In teract ion Curve fo r Short Pr ismat ic Wall F igure 2.2 Mj^  - moment produced by P q applied at edge of kern. The solid bar represents the added moment due to the horizontal deflection, i.e., the left end represents the maximum moment excluding the effect of the vertical load acting on the horizon-tal deflection and the r igh t hand end represents the total maximum moment at failure. The right hand end of the solid bar should be compared to the solid line interaction curve. The l e f t hand end should be compared to the reduced i n t e r a c t i o n curve (broken line) Figure 2.3 P-M Interaction Curve 25. CHAPTER 3 CURRENT ANALYSIS METHODS AND CODE APPROACH It i s intended in t h i s chapter to present a b r i e f d i s c u s s i o n of those concepts which form the b a s i s for the ana lys is and design of p l a i n and r e i n f o r c e d masonry. Masonry s t ruc tures a r e , at present , designed by working s t ress analy-s i s . The concepts involved here are those of the e l a s t i c theory, long app l ied in des igning r e i n f o r c e d concrete elements, but i n t h i s case the ul t imate u n i a x i a l strength of the masonry ( f ) i s used in the a n a l y s i s formulas to r e f l e c t the p roper t i es of masonry. The determination of masonry p r o p e r t i e s i s d iscussed ex tens ive ly i n the next chapter . The d e c i s i o n whether to use p l a i n or r e i n f o r c e d masonry depends on the magnitude of the loads as wel l as minimum requirements es tab l i shed by the codes. 3.1 A n a l y s i s of Walls Under Combined Bending and A x i a l Loads  3.1.1 P l a i n Masonry The load c a r r y i n g capac i ty of a p l a i n masonry wal l subjected to combined bending and a x i a l loads can be determined i f the t e n s i l e and compressive strength of the masonry, as wel l as the s t ress d i s t r i b u t i o n at f a i l u r e , are known. I t i s u s u a l l y assumed that a l i n e a r s t r e s s - s t r a i n r e l a t i o n s h i p e x i s t s up to f a i l u r e , thus , the s t resses can be determined for a given l o a d i n g . Th is type of f a i l u r e can be i l l u s t r a t e d by the fo l lowing examples: 26. (a) For non-cracked sect ions the s t resses at the outer faces w i l l be f = 1 ± *± " (1) A 21 where A = area I = moment of i n e r t i a of the non-cracked s e c t i o n P = app l ied load M = app l ied moment t = th ickness of gross s e c t i o n . I f the t e n s i l e strength of the assemblage i s assumed to be zero , Equa-t i o n (1) i s v a l i d f o r loads app l ied wi thin the kern , i . e . f o r e < e (2) k where 21 e = kern e c c e n t r i c i t y = — ( for a symmetrical sect ion) and k At | ( 3 ) e = M/P 1 (b) For v e r t i c a l loads at e c c e n t r i c i t i e s f a l l i n g outs ide the kern but w i th in the t h i c k n e s s , and fo r assumed zero t e n s i l e s t rength , the s t ress i n a s o l i d sec t ion would be: , ± P r 1 1 . . . max 3 A L l - 2 e / t J K ' (c) In some cases some t e n s i l e strength can be a t t r i b u t e d to the masonry, and the capac i ty of a s o l i d sect ion i s then given by: P = | ^ ( f - f'> (5) 2 max t and « - f (1-1-33 |- [ f ^ f , 2 ] ) o where, u = length of uncracked s e c t i o n b = width of s e c t i o n 27. f = maximum compressive s t ress = a f* max m f ' = t e n s i l e strength = s f t m P = f ' bt o m For masonry with no t e n s i l e strength s = 0 and Equation (6) reduces to Equation (4). S imi la r equations to the ones presented above can be der ived for hollow symmetrical s e c t i o n s . It i s very d i f f i c u l t to f i n d a continuous equation fo r the moment capac i ty of a cracked sec t ion that w i l l be a p p l i -cable to a l l hollow symmetrical sect ions because of the d i s c o n t i n u i t i e s i n the s e c t i o n s . 3.1.2 E l a s t i c A n a l y s i s of Reinforced Masonry Walls P l a i n masonry wal ls are i n some cases inadequate i n s t rength , s t i f f -ness or d u c t i l i t y , therefore the wal ls have to be r e i n f o r c e d . If an e l a s t i c a n a l y s i s i s used then the usual assumptions are made. There are two cases to be considered when analyz ing r e i n f o r c e d masonry w a l l s : (a) the e f f e c t i v e e c c e n t r i c i t y (M/P) i s so small that no t e n s i l e s t ress i s developed i n the s e c t i o n , (b) the e c c e n t r i c i t y i s s u f f i c i e n t to produce a cracked s e c t i o n . 3.1.2(a) Uncracked Sect ion In t h i s case the e f f e c t of the reinforcement i s not s i g n i f i c a n t . There fore , for a s o l i d s e c t i o n , f + f . , max mm P = bt f , where f = (7) av av 2 M = ) ( " ^ , m ± n ) (8) f - f 28. 3.1.2(b) Cracked Sect ion 3 . 1 . 2 ( b . l ) S t e e l i n Compression In t h i s c a s e , as in the previous one the s t e e l can be ignored , then: P = -7 f bpt (9) 4 max 1 rt t n M = - f bpt b - P7 J ( 1 0 ) 4 max L2 ^ 6 J where pt /2 i s the length of the compression zone, and 1.0 < p < 2 .0 . 3 .1 .2(b.2) S t e e l i n Tension In t h i s case from bas ic equ i l ib r ium considera t ions the fo l lowing equa-t ions are obtained: f — = f ( ^ ) , (11) n max p E s t e e l where n = E masonry and fo r the s t e e l p laced at the centre of the s e c t i o n , P = f (f^)b - A f (12) max 2 s s M = f p b t 2 (3-p) (13) max 24 where pt /2 = length of the compression zone, and 0.0 < p < 1.0. 3.1.3 Ul t imate Strength A n a l y s i s ,As i n the case of e l a s t i c a n a l y s i s , the p r i n c i p l e s of u l t imate st rength design f o r r e i n f o r c e d concrete elements can be used to analyze r e i n f o r c e d masonry. In t h i s case the f a m i l i a r f 1 , u n i a x i a l compressive strength of the c concre te , i s rep laced by f^, the u n i a x i a l compressive st rength of the masonry assemblage (the grout and masonry b lock are not considered as two d i f f e r e n t m a t e r i a l s ) . The compression on the masonry i s given by a 29. rectangular s t ress block whose magnitude i s 0.85 f . m 3 .2 Code Design Procedures In the fo l low ing sect ions the procedures fo r des igning masonry wal ls under d i f f e r e n t codes of p r a c t i c e are examined. 3.2.1 Loadbearing Walls 3 .2.1(a) The ACI (Ref. 21) , NCMA (Ref. 22) and UBC (Ref. 23, only fo r  Reinforced Masonry) Codes In the new ACI recommendations, (Ref. 21) the ; equations have been developed to b r i n g non- re in forced and r e i n f o r c e d concrete masonry in to one formula which i s a p p l i c a b l e to both c a s e s . Experience and tes ts ind ica ted that strength of masonry can be s i g n i -f i c a n t l y reduced by poor workmanship and the use of uncont ro l led m a t e r i a l s . According to the ACI and UBC recommendations (Refs. 21 and 23) the designer may use f u l l a l lowable s t r e s s values i f the cons t ruc t ion i s inspec ted , but must use reduced s t ress values i f there i s no i n s p e c t i o n . ACI-531 recom-mends, fo r cases where there i s no i n s p e c t i o n , reduced al lowable s t resses of 2/3 o , 1/2 a and 1/2 T . The UBC code recommends a compr. t e n s . shear reduct ion of the al lowable s t r e s s by one ha l f for a l l the d i f f e r e n t types of s t r e s s . In the ACI p r o v i s i o n s , there i s an increase of o n e - t h i r d i n the a l low-able values for s t resses due to wind or earthquakes combined with dead and l i v e l o a d s , provided the strength of the member i s not l e s s than that required for dead and l i v e loads a lone. The 1968 NCMA (Ref. 22) does not al low t e n s i l e s t resses i n unre in -forced masonry wal ls b u i l t with hollow u n i t s , there fore l i m i t i n g the 30. e c c e n t r i c i t y to be wi th in the kern . For s o l i d unre inforced wal ls and f o r r e i n f o r c e d masonry, cracked sect ions are a l lowed. It i s a lso stated that up to an e c c e n t r i c i t y of t /3 re in fo rced wal ls may be designed assuming uncracked s e c t i o n s . The slenderness reduct ion fac tor C g used by these codes i s given by: «=. - U - fc£)3] <"> where h i s the height of the w a l l . Equation (14) does not d i f f e r e n t i a t e between s o l i d and hollow sec -t i o n s . Other v a r i a b l e s assoc ia ted with slenderness e f f e c t s and not considered .in these design equations a r e : end f i x i t y ( e f f e c t i v e l e n g t h ) , the manner in .which the member i s loaded ( i e . the moment diagram and the r e s u l t a n t d e f l e c t i o n curve) and the r e l a t i o n s h i p between the strength and the modulus of e l a s t i c i t y of masonry. According to Yokel e t a l . (Ref. 10) the j u s t i f i c a t i o n for not cons ider ing some of these v a r i a b l e s may be i n par t a t t r i b u t e d to the fac t that there i s a l i n e a r r e l a t i o n s h i p between f m and E within a c e r t a i n range of masonry s t rength , and the end condi t ions are s i m i l a r fo r most convent ional masonry s t r u c t u r e s . It i s quest ionable whether, with the i n c r e a s i n g use of high strength masonry and of high r i s e masonry c o n s t r u c t i o n , i t i s s t i l l p o s s i b l e to d is regard these v a r i a b l e s without the use of unduly high margins of sa fe ty . The Codes being d iscussed (Refs. 21, 22, 23) recommend that members subjected to combined a x i a l and f l e x u r a l s t resses s h a l l be designed a c c o r d -i n g to the fo l low ing i n t e r a c t i o n equat ion: 3 1 . where f computed a x i a l s t ress a computed compressive bending s t r e s s F a maximum al lowable a x i a l s t r e s s , i n c l u d i n g any s t r e s s increase permitted fo r wind and earthquake and reduced for slenderness e f f e c t by the slenderness c o e f f i c i e n t presented above (Equation 14) F b maximum al lowable f l e x u r a l s t r e s s , i n c l u d i n g any s t r e s s increase permitted for wind and earthquake. Yokel et a l . (Ref. 9) compared tes t r e s u l t s with al lowable s t r e s s i n t e r a c t i o n curves c a l c u l a t e d using the code procedure as given above. The margin of safety was computed in two ways: (a) the r a t i o of average e x p e r i -mental u l t imate loads to al lowable loads for s p e c i f i c load e c c e n t r i c i t i e s and (b) the r a t i o of u l t imate to al lowable moments for s p e c i f i c l e v e l s of v e r t i c a l l o a d s . In both cases , an increase i n the margin of sa fe ty with increas ing slenderness was apparent. The second method, showed margins of safety decreasing with i n c r e a s i n g e c c e n t r i c i t y . The authors (Yokel et a l . ) commented that for e c c e n t r i c i t i e s greater than t /6 the margin of safety against an increase i n l a t e r a l loads ( i . e . moving h o r i z o n t a l l y on the i n t e r a c t i o n diagram) was s i g n i f i c a n t l y smaller than that provided against an increase i n v e r t i c a l l o a d s . In Ref . 10, Yokel et a l . der ived wal l i n t e r a c t i o n curves fo r ul t imate s t resses fo l lowing the p r i n c i p l e s es tab l i shed by the Code (Equation 15). Ult imate s t resses were obtained by m u l t i p l y i n g the code al lowable s t ress by 5, which according to the authors can be considered the a x i a l load margin of sa fe ty (see a lso Ref. 9 ) . Those curves were modi f ied fo r slenderness using Equations (14) and (15). 32. The code curves were compared to tes t r e s u l t s and other i n t e r a c t i o n curves der ived us ing the moment magnif ier method. For the slenderness of the wal ls tested (h / t - 13), the modi f ica t ion of the i n t e r a c t i o n curves was r e l a t i v e l y minor, thus curves for h / t of 30 were constructed to provide a bet ter comparison between the Code approach and the moment magni f ier method. For both slenderness r a t i o s (h / t = 13 and h / t =30 ) i t appeared that the moment mangnif ier curve was more conservat ive than the code curve , except for very small e c c e n t r i c i t i e s . For a small s lenderness r a t i o ( i . e . h / t = 13) the code curve produced a bet ter f i t to the experimental r e s u l t s than the moment magni f ier curve . In the case of greater slenderness (h / t = 30) , although no slender concrete block wal ls were t e s t e d , Yokel et a l . concluded, based on the agreement between observed and pred ic ted strength of more slender b r i c k w a l l s , that the code curve probably overestimates the transverse strength of t ransverse ly loaded slender w a l l s , but i t i s probably conservat ive for the case of eccen t r i c v e r t i c a l l o a d s . 3.2.1(b) Commentaries on par t 4 of the 1975-NBC (Ref. 24) , CSA-CAN3- S304 (Ref. 20) , 1969 SCPI: B u i l d i n g Code Requirements for  Engineered B r i c k Masonry (Ref. 25); UBC Code (Ref. 23, only fo r  Unre inforced Masonry). The UBC and SCPI codes p rescr ibe the use of reduced al lowable s t resses f o r cases where there i s no acceptable i n s p e c t i o n . CSA-CAN3-S304 and the 1975 NBC commentary do not al low uninspected c o n s t r u c t i o n . The UBC and SCPI codes e s t a b l i s h that the al lowable v e r t i c a l load fo r p l a i n sect ions with e < t / 3 i s given by: P = C C f A e s m g (16) 33. where f i s the al lowable a x i a l compression s t r e s s , A the gross cross m g s e c t i o n a l a rea , C g an e c c e n t r i c i t y fac tor and C g a slenderness f a c t o r . CAN3-S304 uses the net area A instead of A . The c o e f f i c i e n t C i s deter -n g e mined by the fo l lowing equat ions: C = 1.0 e 1.3 l , e 1 w , 1 C = — T + 2 ( t - 2 0 ) ( 1 T > e 1+6- 2 1 e 1 e 1 1 c e = 1.95 ( - - - ) + I ( t - ^ ) d - r - ) For e / t < 1/20 1/20 < e / t < 1/6 1/6 < e / t < 1/3 (17) where e = maximum v e r t i c a l e c c e n t r i c i t y e^ = smal ler end e c c e n t r i c i t y e^ = la rger end e c c e n t r i c i t y t = wal l t h i c k n e s s . For some cond i t ions of loading and methods of support , the maximum v e r t i c a l e c c e n t r i c i t y (e) may occur at a l o c a t i o n other than the po int of support . The v e r t i c a l e c c e n t r i c i t y may, t h e r e f o r e , d i f f e r from the eccen-t r i c i t i e s used for the r a t i o of end e c c e n t r i c i t i e s (e^/e^)* The UBC and SPCI s p e c i f i c a t i o n s p rescr ibed a r a t i o e ^ / e 2 e < 3 u a ^- t o fo r members subjected to l a t e r a l loads higher than 10 p s f . No l a t e r a l loads e f f e c t s are taken in to cons idera t ion by CAN3-S304. The slenderness fac tor C i s given by: s C = 1.2 - [5.75 + (1.5 + —) 2 ] < 1.0 s 300 e 2 :i8) This fac tor ( s i m i l a r to the one presented i n sec t ion (a) equation 14) does not e x p l i c i t l y take in to cons idera t ion some of the v a r i a b l e s which i n f l u -ence the slenderness e f f e c t , e . g . : the load l e v e l and load ing c o n d i t i o n s , 34. v a r i a t i o n in the moment of i n e r t i a due to c r a c k i n g , the end c o n d i t i o n s , and the f ' / E r a t i o , m Loads app l ied at e c c e n t r i c i t i e s greater than t /3 are l i m i t e d by the al lowable f l e x u r a l t e n s i l e s t r e s s . When t h i s al lowable s t r e s s i s exceeded the sec t ion has to be r e i n f o r c e d (Refs. 20, 23, 25). The recommendations presented in the Commentary to Part 4 of the 1975 NBC (Ref. 24) show a s i m i l a r approach as the one stated above except that A g i s replaced by the area A n i n Equation (16). The slenderness and e c c e n t r i c i t y c o e f f i c i e n t s are a lso s l i g h t l y d i f f e r e n t and given by: C = 1 - CL (7 - 5) s b t where 0.003 ( e , / e ) z + 0.012 (e / e ) + 0.025 1 2 1 2 (19) (20) and C = 1.0 C = C = C = 4 ( 1 " 2?> f o r e / t < 1/20 fo r 1/20 < e / t < 1/6 fo r 1/6 < e / t < 1/3 fo r 1/3 < e / t (21) For e / t > 1/3, f i n Equation (16) i s replaced by the al lowable ten -m s i l e s t ress f. For r e i n f o r c e d masonry, use i s made of the same equations as for unre inforced masonry, but in t h i s case the 1975 NBC Commentary increases the e c c e n t r i c i t y f o r which bending i s ignored from t /20 to t / 1 0 . For e c c e n t r i c i t i e s greater than t /3 or a value producing t e n s i l e s t resses i n the s t e e l b a r s , the a x i a l load capaci ty P i s to be taken as the short wal l 35. c a p a c i t y , c a l c u l a t e d on the b a s i s of a transformed sec t ion and a l i n e a r s t ress d i s t r i b u t i o n , reduced by the slenderness c o e f f i c i e n t C s def ined above. For r e i n f o r c e d masonry, the SPCI and CAN3-S304 Codes present a s im i la r approach as the one descr ibed above, except that the d e f i n i t i o n of the c o e f f i c i e n t C i s the same as that of p l a i n masony ( for e / t < 1/3) . e A l l the above codes approach the case of b i a x i a l bending i n the same way as they do fo r u n i a x i a l bending where e / t i s rep laced by (be + te ) / b t . t b In Ref. 10, Yokel et a l . compared the r e s u l t s of t e s t s on l a t e r a l l y loaded b r i c k wa l ls with an i n t e r a c t i o n curve der ived us ing the moment magnif ier method (using C = 1.0, k = 0 .8 ) , and a lso with an i n t e r a c t i o n m curve based on the p re l im inary 1969 SCPI Standard where the ul t imate loads were taken as C C f A . The code curve was developed for eccen t r i c e s m g v e r t i c a l load with e ^ / e 2 = - n ' 4 , which assumes p a r t i a l f i x i t y at one end and a pinned cond i t ion at the other end. Th is curve p red ic ted a x i a l load c a p a c i t i e s that were i n good agreement with tes t r e s u l t s and the moment magnif ier curve except fo r small va lues of P, where there was a cons ider -able d i f f e r e n c e i n the moment c a p a c i t i e s due to the d i f f e r e n c e s i n load ing and end condi t ions between Y o k e l ' s t e s t program and those assumed by the 1969 SPCI code (see Figure 3 .1 ) . When Yokel et a l . changed to the condi t ions used i n the SPCI code ( i . e . C = 0.5 and k = 0 .8 , see Figure m 3 .1 ) , the moment magni f ier method for e c c e n t r i c v e r t i c a l loads agreed approximately with the SPCI curve . The o r i g i n a l SPCI recommendations were developed on the b a s i s of t e s t s 36. on wal ls loaded with eccen t r i c v e r t i c a l loads o n l y , and so they were not n e c e s s a r i l y appropriate for the case where the wal l i s a l s o loaded l a t e r -a l l y . The case of l a t e r a l loading was recognized i n the 1969 SPCI Standard (see Ref . 25, Sect ion 4 .7 .6 .2 (5 ) (c ) ) as a r e s u l t of the i n v e s t i g a t i o n of Yokel e t a l . (Ref. 10). Based on tes t da ta , the new SPCI code recommends that f o r wal ls subjected to combined v e r t i c a l and transverse l o a d i n g , C^ should be c a l c u l a t e d using an e . / e value of +1.0, but C and the maximum slenderness r a t i o should be c a l c u l a t e d on the b a s i s of the actua l value o f w Considera t ion to l a t e r a l loads i s a lso given i n the UBC Code (same recommendations as SPCI Code), but not in the Canadian Codes (Refs. 20 and 21) . 3 .2 .1(c) Minimum Thickness f o r Loadbearing Walls The ACI Code (Ref. 21) e s t a b l i s h e s a minimum thickness for masonry wal ls of 1/36 times the l e a s t d is tance between l a t e r a l supports e i ther h o r i z o n t a l or v e r t i c a l . For non- re in forced bear ing wal ls the r a t i o i s increased to 1/20. The UBC Code l i m i t s the s i z e s of var ious masonry wal ls i n the form of a s p e c i f i e d minimum wal l th ickness ( e . g . : 6 i n . for r e i n f o r c e d walls) and a maximum unsupported height to th ickness r a t i o ( e . g . : h / t = 25 fo r re in fo rced hollow masonry). Amrhein (Ref. 26) expla ins that the l i m i t i n g value of h / t = 25 f o r r e i n f o r c e d grouted or hollow masonry, was es tab l i shed in order to l i m i t the tens ion on a mortar j o i n t to 50 p s i when the wal l i s subjected to a wind load of 15 p s f , with no account taken of the se l f -we ight of the w a l l . The same explanat ion i s presented by Schneider and Dickey (Ref. 37. 27) . According to the UBC recommendations, the h / t r a t i o may be increased and the minimum th ickness decreased "when data i s submitted which j u s t i f i e s a reduct ion i n the requirements s p e c i f i e d " , fo r example, making some p r o v i -s ion fo r end r e s t r a i n t s of the w a l l . In the Canadian Codes (CAN3-S304 and 1975 NBC) minimum requirements are given as p a r t . o f the "Empi r ica l Rules for P l a i n Masonry Design not Based on Engineer ing A n a l y s i s " . For unre inforced masonry i t i s recom-mended: (1) h / t < 18 for s o l i d masonry made of hollow u n i t s and (2) h / t < 36 f o r p a r t i o n w a l l s . A l l the codes ( fo r UBC t h i s app l ies only to unre in forced masonry) set an upper bound for the slenderness r a t i o of a load bear ing w a l l : h / t should not exceed 10(3 - e ^ / e 2 ^ where (^/^^ > 0 f o r s i n 9 " l e curvature and ( e ^ / e ^ = 0 when e.^  and e 2 are zero . The UBC Code (Ref. 23) and the SPCI Standard (Ref. 25) s t i p u l a t e that when the wal ls meet a l l other r e q u i r e -ments, the designer may present a wr i t ten j u s t i f i c a t i o n i n order to d i s r e -gard the l i m i t on the s lenderness . 3.2.1(d) Minimum Reinforcement for Loadbearing Walls In the p lanning of a b u i l d i n g , i t i s e s s e n t i a l to determine whether the masonry should be p l a i n or r e i n f o r c e d . Part 4 of the NBC (Ref. 28) d i c t a t e s that i n seismic zones 2 and 3, masonry must be r e i n f o r c e d i n l o a d -bear ing and l a t e r a l l o a d - r e s i s t i n g s t r u c t u r a l elements, i n wal ls enc los ing e levator shaf ts and s ta i rways , i n ex te r io r c ladd ing and i n c e r t a i n categor-i e s of p a r t i t i o n s . Only i n zones 0 and 1 does the designer have the opt ion to s e l e c t e i t h e r p l a i n or r e i n f o r c e d masonry. Thus, i n CAN3-S304, s ince a l l r e i n f o r c e d masonry i s subjected to design by engineer ing a n a l y s i s the use o f the convent ional r u l e s f o r p l a i n masonry (Sect ion 5) i s r e s t r i c t e d 38. to masonry i n b u i l d i n g s in zones 0 and 1. A l l codes d iscussed in t h i s chapter requi re a t o t a l minimum s t e e l area of 0.002 times the gross cross s e c t i o n a l area of the wal l and that not l e s s than a t h i r d of i t be p laced e i ther h o r i z o n t a l l y or v e r t i c a l l y . The maximum spacing between bars i s set by the UBC Code at 4 f ee t (1.22 m). The 1969 SPCI, 1975 NBC Commentary and S304 have the same upper l i m i t but add that i t should not exceed 6 times the th ickness of the w a l l . According to ACI-531 the spacing between bars s h a l l not exceed 12 times the wal l th ickness nor 8 feet (2.44 m) on centers i n each d i r e c t i o n . The maxi-mum h o r i z o n t a l s t e e l spacing i s reduced to 4 fee t on centers when the wal l i s f u l l y grouted. The ACI standard a l s o requ i res that the bound area between v e r t i c a l and hor i zon ta l bars should not exceed 32 square feet (~3 m 2 ) , and when the wal l i s subjected to large loadings or movements i t i s recommended that the s i ze of the bound area be reduced to 20 square feet (-1.9 m 2 ) . CAN3-S304 and the UBC Code, a l low j o i n t reinforcement (wire r e i n f o r c e -ment) to be considered as requi red h o r i z o n t a l re inforcement . Amrhein (Ref. 26) expla ins that many of the requirements that were used for r e i n f o r c e d concrete were rev ised and app l ied to r e i n f o r c e d masonry. Minimum r e i n f o r c i n g s t e e l i s p laced i n concrete to compensate for the shr inkage, moisture changes, and temperature s t resses that develop not on ly dur ing the l i f e o f the s t r u c t u r e , but a l s o dur ing the i n i t i a l p l a c i n g and cur ing of concre te . In masonry the u n i t s have already experienced most of the shrinkage that w i l l occur before the wal l i s const ruc ted . Mortar and grout do shr ink but comprise only about one-ha l f of the volume of the wal l even i f f u l l y grouted. There fore , only h a l f or l e s s of the mater ia l that shr inks i s used and for t h i s reason the codes u s u a l l y requi re only 39. about ha l f as much minimum reinforcement i n masonry as i n concre te . 3.2.2 Non-Loadbearing Walls Non- loadbearing wal ls are wal ls which support no v e r t i c a l load other than t h e i r own weight. Th is c l a s i f i c a t i o n may, t h e r e f o r e , inc lude earth r e t a i n i n g wal ls and wal ls of hydrau l ic s t r u c t u r e s , as we l l as i n t e r i o r and ex te r io r non- loadbearing wal ls of b u i l d i n g s . When due to the magnitude of the appl ied load or to the method of support , non- loadbear ing wal ls requi re r e i n f o r c i n g s t e e l , they should be designed i n accordance to the Codes' gu ide l ines fo r the design o f f l e x u r a l members. 3.2.2(a) Minimum Thickness For Non-Loadbearing Walls The 1969 SCPI standard provides l a t e r a l slenderness gu ide l ines for non- loadbear ing, non- re in forced b r i c k masonry w a l l s , subjected to var ious l a t e r a l l o a d s . The recommendations are based on the assumption that the wal l i s simply supported and contains no openings or other i n t e r r u p t i o n s . The a n a l y s i s neglects the weight of the wal l and the a l lowable s t resses are increased by 33.3% for wind. Prov is ions are made a lso for wal ls supported i n two d i r e c t i o n s ; i n that case the d is tance between supports can be increased but the sum of the hor i zon ta l and V e r t i c a l span cannot exceed 3 times the al lowable d is tance permitted for support i n one d i r e c t i o n . The UBC Code a lso gives a r a t i o of unsupported height or length to th ickness with a minimum thickness requirement of 2 inches (5 cm). 3.2.2(b) Minimum Reinforcement fo r Non-Loadbearing Walls The UBC Standard and 1975 NBC recommendations do not d i f f e r e n t i a t e between loadbear ing wal ls and non-loadbearing wal ls when e s t a b l i s h i n g the minimum s t e e l requirements. In the "Recommended P r a c t i c e for Engineered B r i c k Masonry" of the SCPI Standards ( Ref . 25) i t i s requ i red that non-40. loadbear ing wal ls should be designed i n accordance with the f l e x u r a l design of r e i n f o r c e d b r i c k masonry. The minimum.steel r a t i o f o r f l e x u r a l members should be not l e s s than 80/ f , unless the reinforcement provided at every y s e c t i o n i s a t l e a s t o n e - t h i r d greater than that requ i red by a n a l y s i s , where f y i s the s t e e l y i e l d s t r e s s i n p s i . CSA Standard CAN3-S304-M78 estab-l i s h e s that for non- loadbear ing wal ls when reinforcement i s r e q u i r e d , i t should be provided i n one or more d i r e c t i o n s with r e i n f o r c i n g s t e e l having a minimum area of 0.0005 A i n seismic zone 0, 1 and 2 and 0.001 A in g g seismic zone 3. The maximum spacing for one way reinforcement i s 16 inches (400 mm) whereas i f a t l e a s t o n e - t h i r d of the s t e e l i s p laced i n a second d i r e c t i o n the maximum spacing can be increased to 4 feet (1.22 m) Th is reduces the minimum requirements f o r non- loadbear ing wal ls i n compar-ison with the requirements es tab l i shed in the NBC 1975 and the UBC codes; i t has a more r a t i o n a l approach and takes in to account the d i f f e r e n t s e i s -mic zones. The designer must ensure that the reinforcement i s adequate to r e s i s t the design seismic f o r c e s . J o i n t reinforcement can be used to meet the minimum requirements. (a) Yokel et a l . (Ref. 10) I' pin p a r t i a l f i x i t y (b) 1969 SCPI f pm \ s/ss/s. - p a r t i a l f i x i t y Note: Case (a) when using the moment magni f ier method w i l l have Cm = 1.0 and k = 0.8. Case (b) w i l l have Cm = 0.5 and k = 0.8. End and Loading Condit ions Assumed by (a) Yoke l ' s Ser ies of Tests i n Ref. 10 and (b) 1969 SCPI Code Figure 3.1 42. CHAPTER 4 MATERIAL PROPERTIES In t h i s chapter the p roper t i es of the i n d i v i d u a l mater ia ls and o f composite masonry assemblages are studied both t h e o r e t i c a l l y and e x p e r i -mental ly . A l l the mater ia ls used in the const ruc t ion of the t e s t specimens at the U n i v e r s i t y o f B r i t i s h Columbia are commercial ly a v a i l a b l e and t y p i -c a l of those commonly used i n concrete masonry c o n s t r u c t i o n . 4.1 Concrete Block Un i ts Concrete b locks are c l a s s i f i e d as hollow or s o l i d u n i t s . A hollow u n i t i s def ined as one i n which the net area i s l e s s than 75% of the gross cross s e c t i o n a l area . The net cross s e c t i o n a l area of most concrete un i ts ranges from 50% to 70% depending on the un i t width , face s h e l l , web t h i c k -ness , and core c o n f i g u r a t i o n . Because of t h e i r l i g h t e r weight and e a s i e r handl ing , hollow u n i t s are more popular than s o l i d u n i t s . For s t r u c t u r a l reasons some standards require a minimum face s h e l l and web th ickness ( e . g . : ASTM: C-90, CSA 165). Concrete u n i t dimensions are usua l ly based on modules of 4 or 8 i n . (100 or 200 mm). From common usage the 3/8 i n . (~10 mm) t h i c k mortar j o i n t has become standard, therefore the ex te r io r dimensions of modular un i ts are reduced by the th ickness of one mortar j o i n t . The nominal block s i z e that dominates the industry i s 8 x 8 x 16 i n . (200 x 200 x 400 mm). The major i ty of concrete b lock u n i t s produced i n Canada are c l a s s e d as Concrete Masonry Uni ts under CSA A165.1 (Ref. 29) , and some are a lso manu-fac tured to conform the requirements of the American Soc ie ty fo r Tes t ing and M a t e r i a l s : ASTM C-90 (Ref. 30) . CSA Standard A165.1 "Concrete Masonry 43. Units"/ c l a s s i f i e s masonry u n i t s , except concrete b r i c k s , by t h e i r p h y s i c a l properties using the four-facet system. For instance, H/1000/C/0 i s a hollow unit with a strength of 1000 p s i (average of 5 u n i t s ) , a density of le s s than 105 pcf and a undefined moisture content at the time of shipment. The standard does not f i x the weight, colour, surface texture, f i r e r esistance, thermal transmission or aco u s t i c a l properties of the blocks. In the CSA s p e c i f i c a t i o n s c e r t a i n type of blocks are excluded for exterior use. ASTM s p e c i f i c a t i o n s c l a s s i f y concrete masonry units according to grade and type. The grade describes the intended use of the concrete masonry uni t s , while the type r e f e r s to the moisture c o n t r o l of the u n i t : type I moisture c o n t r o l l e d , type II non-moisture c o n t r o l l e d . S i m i l a r l y to the CSA standard, the ASTM does not e s t a b l i s h the required weight, colour, surface texture, f i r e r e s istance, thermal transmission or ac c o u s t i c a l properties of the u n i t s . 4.1.1 Compressive Strength of Masonry Blocks The determination of the block compressive strength i s one of the most d i f f i c u l t aspects of t e s t i n g masonry units. I t i s also very important because many masonry codes allow the use of a formula to c a l c u l a t e the design value of masonry compressive strength ( f ) , and t h i s formula uses m the u n i a x i a l compressive strength of the masonry u n i t . Normally the blocks are capped before t e s t i n g and are subjected to compressive stress i n a standard t e s t i n g machine, with the capping d i r e c t l y 44. against the s t e e l surface of the t e s t i n g machine. J . J . Roberts (Ref. 31) reported a program i n which one of the aims was to compare var ious methods of capping; s i n g l e block specimens cons is ted of 4 types: (1) mortar-capped blocks tested wet (2) " " " " dry (3) board-capped blocks tested wet (4) " » " « d r y # Board-capping was used i n the tes ts because i t has severa l advantages over mortar capping: i t i s e a s i e r , quicker and cheaper. A f te r using f i v e types of f i b r e - b o a r d the authors concluded that the type of f i b r e - b o a r d had l i t t l e e f f e c t upon the mean ind ica ted block s t rength : four types gave s i m i l a r l y cons is ten t r e s u l t s but one case y i e l d e d a somewhat l a rger c o e f f i -c i e n t of v a r i a t i o n . Board-capped specimens produced a lower ind ica ted strength than mortar-capped specimens. In 1980, W. Ridinger et a l . (Ref. 32) presented the r e s u l t s of an i n v e s t i g a t i o n that was p r i m a r i l y designed to evaluate the in f luence of capping, loading condi t ions and the in f luence of reduced in te r face f r i c t i o n upon the r e s u l t s of u n i a x i a l compressive t e s t s of hollow c l a y u n i t s . The authors t r i e d to reduce the i n t e r f a c e f r i c t i o n by using an i n t e r -face assembly c o n s i s t i n g of two layers of polyethelene p l a s t i c (4 m i l ) , separated by a t h i n layer of high v i s c o s i t y l u b r i c a n t . In order to consider the in f luence of the loaded a rea , some specimens were capped only i n the face s h e l l and i n others the capping was app l ied to the net area of the u n i t (current p r a c t i c e ) . For specimens loaded over the e n t i r e net area with f u l l i n t e r f a c e l a t e r a l r e s t r a i n t , the f a m i l i a r pyramidal mode of 45. f a i l u r e , t y p i c a l of shear f a i l u r e , was observed. The un i ts loaded on the face s h e l l s o n l y , with f u l l i n te r face r e s t r a i n t , showed evidence of h o r i z o n t a l compressive forces wi thin the cross-webs, together with s p a l l i n g of the outer por t ions of the face s h e l l s . Uni ts tes ted with reduced i n t e r f a c e f r i c t i o n seemed to expand more f r e e l y at t h e i r upper and lower boundar ies , r e s u l t i n g i n v e r t i c a l tens ion c r a c k i n g . For face s h e l l loaded specimens with reduced in te r face f r i c t i o n , a p h y s i c a l separat ion of the face s h e l l s from the cross-webs was a lso observed, as wel l as l o c a l i z e d s p l i t t i n g wi th in the face s h e l l s themselves. The v i s u a l resemblance of the f a i l e d specimens to the prism and wal l f a i l u r e s suggests that t h i s t es t may be a more accurate i n d i c a t i o n of un i t compressive capac i ty than the current standard t e s t . The un i t compressive strength based on these tes ts was between 73 and 78% of the u n i t compressive strength based on the normal t e s t where the ends are res t ra ined somewhat. Since the l a t t e r t e s t i s the current indust ry standard, present compressive strength t e s t s of hollow c l a y masonry u n i t s may give erroneously high impressions of i n - p l a c e compressive c a p a c i t y . 4 .1 .1(a) Block Compressive Strength Tests at the U n i v e r s i t y of B r i t i s h  Columbia. The bas ic u n i t s used for cons t ruc t ing a l l t e s t specimen wal ls were the 8 x 8 x 16 i n . (200 x 200 x 400 mm) s t re tcher b l o c k , the 8 x 8 x 16 i n . end b l o c k , the 8 x 8 x 16 i n . knock out b l o c k , and the 8 x 8 x 8 i n . h a l f b l o c k . A l l the un i ts were manufactured by Ocean Const ruct ion Suppl ies L t d . of Vancouver. Block compressive strength t e s t specimens were se lec ted from 8 x 8 x 16 i n . s t recher and end b l o c k s . The tes ts were made i n conformance with 46. A165.1 Concrete Masonry Uni ts as required by CSA Standard CAN3-S304-M78 (Ref. 20). The bear ing surfaces of the u n i t s were capped with Hydrostone gypsum cement. The r e s u l t s are l i s t e d i n Table 4 . 1 . According to CSA.CAN3-S304 the compressive strength s h a l l be obtained by subt rac t ing one and a ha l f times the standard dev ia t ion from the average compressive s t rength . In our case: on the net area: a , , 3520 p s i (24.3 MPa) block and on the gross area: ° b l o c k = * 8 6 5 p s i < 1 2 - 9 MPa) 4.1.2 Block T e n s i l e Tests One of the major d i f f i c u l t i e s faced by researchers i n t h i s subject i s the development of t es t techniques that w i l l determine p roper t i es adequate fo r the a n a l y s i s of l a rger elements. Many d i f f e r e n t t e s t i n g techniques have been proposed and used i n var ious i n v e s t i g a t i o n s , with a consequent lack o f c o r r e l a t i o n between them. I n d i r e c t t e s t s are c a r r i e d out by support ing the u n i t as a beam on two r o l l e r supports and loading i t at mid-span i n order to determine the modu-lus of rupture . D i r e c t t e n s i l e t e s t s cons is ted of g lu ing p l a t e s to the u n i t ends or clamping the ends with s p e c i a l g r ip devices and then p u l l i n g apart the u n i t . Extreme care i s requ i red when performing d i r e c t t e n s i l e strength t e s t s to minimize e c c e n t r i c i t y of the l o a d i n g . 4.1.2(a) Block T e n s i l e Tests at the U n i v e r s i t y of B r i t i s h Columbia At the U n i v e r s i t y of B r i t i s h Columbia t e n s i l e s t rength of concrete blocks was determined by d i r e c t tension t e s t s . The experimental specimens were the teeth of knock-out b l o c k s , clamped at t h e i r ends with a g r i p device and p u l l e d i n the d i r e c t i o n of the long dimension. 47. Test r e s u l t s are shown i n Table 4.II. 4.2 Mortar Mortar c o n s i s t s of a p l a s t i c , workable mixture of cement, sand, water and l ime . Other admixtures might be added because of a r c h i t e c t u r a l or engineer ing requirements. Mortar f o r concrete masonry should be designed to (1) j o i n the blocks in to an i n t e g r a l s t r u c t u r e , (2) sea l i r r e g u l a r i t i e s of the masonry b l o c k s , p rov id ing a weathert ight wal l and prevent ing pene-t r a t i o n of wind and water in to and through the w a l l , (3) bond with s t e e l j o i n t re inforcement , metal t i e s and anchor b o l t s , so that they become an i n t e g r a l par t of the masonry assemblage, and (4) compensate for s i ze v a r i a -t ions i n the un i ts by p rov id ing a bed to accommodate to lerances of the b l o c k s . Good mortar i s necessary for good workmanship and proper s t r u c t u r a l performance of masonry c o n s t r u c t i o n . The main p roper t i es of mortar a re : (a) Workab i l i t y . The workab i l i t y of the f resh mortar must be such that the mason can f i l l a l l the j o i n t s e a s i l y ; i t should ease p l a c i n g of the u n i t without subsequent s h i f t i n g due to i t s weight or the weight of success ive courses . (b) Water r e t e n t i v i t y . Th is property i s r e l a t e d to w o r k a b i l i t y . Rapid l o s s of water might cause the mortar to s t i f f e n too f a s t prevent ing the achievement of good bond and water - t ight j o i n t s . The p roper t i es of the b l o c k s , e s p e c i a l l y the s u c t i o n , p lay an important ro le i n water r e t e n t i v i t y . (c) The Rate of Hardening. The ra te of hardening of mortar due to hydra-t i o n , i f too r a p i d , may reduce the workab i l i t y and bond s t rength; very slow hardening may cause the mortar to f low. 48. (d) Bond. Th is property i s in f luenced by: (1) extent of bond or degree of contact of the mortar with the masonry u n i t s , and (2) t e n s i l e bond s t rength , which i s both a chemical and mechanical a c t i o n . Bond i s a f fec ted by: the mortar components and t h e i r pro-p o r t i o n s , c h a r a c t e r i s t i c s of the masonry u n i t s , workmanship, and cur ing c o n d i t i o n s . The bond strength of the mortar increases as the cement content increases and a lso as the water content increases (though mortar compressive strength decreases as the water cement r a t i o i n c r e a s e s ) . (e) Compressive St rength . The compressive st rength of a masonry assemb-lage may be increased with a stronger mortar, but t h i s increase i s not p r o p o r t i o n a l to the increase i n the compressive strength of the mortar. It has been found experimental ly that an increase of 130% i n the mortar compressive strength r e s u l t s i n only a 10% increase in the compressive strength of concrete masonry w a l l s . Compressive strength of mortar increases with an increase i n cement content and decreases with an increase i n a i r , l ime or water content . Compressive strength measurement invo lves c a s t i n g , cur ing and t e s t i n g 2 - i n . (50 mm) cubes i n compression (CSA A179M, CSA A8, ASTM C270). The current s p e c i f i c a t i o n s fo r mortars (ASTM C270, CSA A179M) c l a s s i f y f i v e types of mortars: M, S, N, O and K. Mortar types are iden-t i f i e d by property or propor t ion s p e c i f i c a t i o n , but not both . Mortar type c l a s s i f i c a t i o n under the property s p e c i f i c a t i o n s i s dependent s o l e l y on the compressive s t rength . The propor t ion s p e c i f i c a t i o n i d e n t i f i e s mortar type through var ious combinations of por t land cement with masonry cement. When not otherwise s p e c i f i e d the propor t ion s p e c i f i c a t i o n governs. 49. In Canada the s e l e c t i o n of mortars depends on whether or not the masonry i s design using engineer ing a n a l y s i s . When t h i s approach i s used, types M, S or N are requ i red (CSA-CAN3-S304). 4.2.1 Mortar Tests at the U n i v e r s i t y of B r i t i s h Columbia The mortar used at the U n i v e r s i t y of B r i t s i h Columbia was propor -t ioned to meet the proport ions by volume s p e c i f i c a t i o n corresponding to Port land cement-lime type S. A c o r r e c t i o n was made in the aggregate por -t i o n (sand) i n order to compensate for bu lk ing as requi red i n c lause 9.3 of the CSA Standard A179M (Ref. 33) . The mortar contained by volume 1 par t of type 10 por t land cement, 1/2 par t of hydrated l ime , and between 4 1/2 and 5 par ts of masonry sand ( f ine sand). The mortar was mixed i n a r o t a t i n g drum concrete mixer . Retempering was permitted but no mortar was used that was more than 2 hours o l d . 4 .2.1(a) Compression Tests Mortar cubes 2 x 2 x 2 i n . (50 x 50 x 50 mm) were cast and cured i n accordance with CSA-Standard A179M (Ref. 33), and CSA-Standard A8 (Ref. 34). Later i n our experimental work, the fo l lowing changes were i n t r o -duced i n the cubes c a s t i n g procedure as descr ibed i n CSA-A8 i n order to reproduce more accura te ly the cond i t ions of the mortar i n the j o i n t s : (a) The sample mortar was taken from the batch and spread on a wooden board where i t res ted for a couple of minutes. (b) I t was then spread on the face of a concrete b l o c k , i n an attempt to reproduce the water suct ion by the masonry un i ts i n the w a l l s . (c) A f te r a minute i t was then p laced i n the moulds. 50. The r e s u l t s are summarized on Table 4. I I I . It can be seen that although the mater ia ls were proport ioned i n order to obtain a por t land cement-lime mortar of type S the compression strength never reached the requi red minimum compressive strength of 1800 p s i (12.4 MPa); i n other words we obtained a strong type N mortar . Two f a c t o r s that might have in f luenced t h i s strength reduct ion are : (a) the increased pro -por t ion of masonry sand (a f ter adjustment for bulking) although i n one case the mortar was proport ioned with 4 1/2 p a r t s of sand (b) the water cement r a t i o . The range of strengths was wide, but there was consis tency i n the values obtained for mortar-cubes sampled on the same day. There was no data a v a i l a b l e to expla in the v a r i a t i o n i n measured s t rengths , although the water cement r a t i o and the change i n experimental procedure could be mentioned as two of the p o s s i b l e f a c t o r s causing that v a r i a t i o n . 4 .2.1(b) Tens i l e -Bond Strength Test at the U n i v e r s i t y of B r i t i s h  Columbia The bond strength i s u s u a l l y assessed on the b a s i s of compressive strength values obtained from 2- inch cubes. At U . B . C . the t e n s i l e - b o n d strength was measured by a d i r e c t tension t e s t , on samples made out of two teeth of knock-out b locks jo ined together by the mortar. The samples were cast at the time of the wal l c o n s t r u c t i o n . They were l i n e d up and l eve led on a wooden board where they kept u n t i l they were t e s t e d . During that t ime, they were covered with a polyethylene p l a s t i c sheet i n order to reproduce the moisture cond i t ions of the mortar i n the j o i n t s (mainly i n the inner par t of the j o i n t ) . In order to t e s t the specimens a s p e c i a l g r i p device that clamped the teeth at t h e i r ends was developed (and used a l s o f o r the tens ion t e s t s o f concrete b l o c k s ) . S p e c i a l care was taken i n order to minimize the bending of the specimens at the time of the t e s t s . 5 1 . Test r e s u l t s are shown on Table 4. IV. It i s obvious that there i s a l o t of sca t te r or v a r i a b i l i t y i n the r e s u l t s , some of which can be a t t r i b u t e d to (a) the rate of load ing (some-times i t was too f a s t , and i t was a l l over before the speed could be adjusted) (b) e c c e n t r i c i t y of the load r e s u l t i n g i n a premature f a i l u r e . 4.3 Grout Grout i s a high-slump concrete made with small aggregate. It must be f l u i d enough to f i l l a l l voids without segregation and completely encase the re inforcement . I t s main funct ion i s to bond (a) the wythes together i n a composite wal l (b) the r e i n f o r c i n g s tee l to the masonry. Grout fo r use i n concrete masonry wal ls s h a l l comply with r e q u i r e -ments of CSA Standard A179M (Ref. 33) [Similar to ASTM C476 (Ref. 35) ] . Admixtures are used i n cases where i t i s necessary to reduce ea r ly water l o s s by absorpt ion by the masonry b l o c k s , to promote bonding of the grout to a l l i n t e r i o r surfaces of the u n i t s , and to produce a s l i g h t expansion s u f f i c i e n t to help ensure complete f i l l i n g of the c a v i t i e s . The excessive water i n the f l u i d grout i s absorbed by the masonry b l o c k s , thus reducing the apparently high water-cement r a t i o . The consis tency of the grout i s measured using a slump t e s t (ASTM C143). Slump i s not s p e c i f i e d i n most codes, however, when slump i s measured us ing ASTM C143-Standard Method of Test for Slump of Por t land Cement Concrete , the des i red slump i s 8 i n . (20 mm) fo r un i ts with low absorpt ion and up to 10 i n . (250 mm) for un i ts with high absorpt ion ( e . g . ; the Commentary-ACI-531R-79 recommends a minimum of 8 i n . ) . Some b u i l d i n g codes ( e . g . UBC code Ref. 23) requ i re a minimum com-press ive st rength of 2000 p s i (1.4 MPa) f o r grout at 28 days, when 52. tested according to UBC-No. 24-23 (or CSA-A179M). The Commentary on ACI-531 (Ref. 21) requi res that the grout compressive strength be at l e a s t equal to that of the requi red strength of the masonry to ensure a higher working r e l a t i o n s h i p (and so impl ies c lause 4 .3 .3 .7 i n CSA-S304). 4.3.1 Grout Compression Tests at the U n i v e r s i t y of B r i t i s h Columbia The compressive strength of the grout used i n grout ing the r e i n f o r c e d wa l l s tes ted at U . B . C . was evaluated experimental ly by t e s t i n g specimens prepared in accordance with the general gu ide l ines e s t a b l i s h e d by CSA Standard A179M, except that they were not cured with an impermeable sheet i n order to r e t a i n t h e i r moisture content . A change was a lso introduced regarding the capping of the specimens: instead of being capped 48 hours a f te r t h e i r c a s t i n g , they were capped on the same day of the compression t e s t , as f o l l o w s : a) a l l the specimens were a i r d r ied for 1 hour b) they were then capped with s u l f u r c) they were then returned to the moisture room and tes ted wet a f t e r 2 hours. The capping was done acording to ASTM-C617. The r e s u l t s are summarized on Table 4 .V . 4.4 Determination of the compressive Strength of Masonry ( f ' ) « m The compressive strength of masonry i s the most important parameter i n the design of masonry s t r u c t u r e s . In order to e s t a b l i s h the ul t imate design strength of the masonry assemblage ( f ) , the UBC Code (Ref. 23) and m CSA-S304 (Ref. 20) al low two methods: (1) an estimate based on the masonry u n i t s t rength and mortar type , which i s purposely conserva t i ve , 53. (2) prism t e s t s , which are r e l a t i v e l y d i f f i c u l t to perform but w i l l gener-a l l y provide the designer with higher al lowable s t r e s s . 4 .4 .1 Pr ism Tests When f* i s to be e s t a b l i s h e d by t e s t s , the specimens c o n s i s t o f m prisms made from the wal l m a t e r i a l s . The moisture content , the c o n s i s t -ency at the time of l a y i n g , the mortar j o i n t th ickness as wel l as the workmanship, should be the same as for the ac tua l w a l l s . The un i t com-press ive s t ress for each specimen i s obtained by d i v i d i n g the u l t imate load by the net area of the b locks and m u l t i p l y i n g the r e s u l t by a c o r r e c -t i o n fac tor depending on the th ickness to height r a t i o of the pr ism. (See Table 1 i n CSA-S304, Ref . 20). The code s p e c i f i c a t i o n s (Ref. 20, and 23), require at l e a s t f i v e specimens to be tested and the compressive strength to be taken as the average f a i l u r e s t ress l e s s one and a h a l f standard d e v i a t i o n s . 4 .4.1(a) V a r i a b l e s In f luenc ing Prism Compressive Strength The compressive strength of an assemblage u s u a l l y l i e s between the compressive strength of the mortar and that of the masonry b l o c k . The modulus of e l a s t i c i t y of the mortar i s u s u a l l y smal ler than that of the masonry u n i t , r e s u l t i n g i n l a rger l a t e r a l deformations in the mor-tar than i n the b l o c k s . I f the P o i s s o n ' s r a t i o of the mortar i s greater than of the b l o c k s , t h i s w i l l r e s u l t in even bigger d i f f e r e n c e s i n the l a t e r a l s t r a i n s . Because of f r i c t i o n and bond, the b locks r e s t r a i n the l a t e r a l expansion of the mortar , producing tension in the b locks and some-times r e s u l t i n g i n t e n s i l e f a i l u r e before compressive f a i l u r e . 54. The in f luence of the mortar composition on the prism s t rength , has been found by some i n v e s t i g a t o r s to be of some s i g n i f i c a n c e . Ref. 36 notes a reduct ion i n prism strength of more than ha l f as one goes from type M to Type 0 mortar which i s a reduct ion i n mortar strength of a f a c -tor of 12. The strength of prisms b u i l t with high bond mortar was found to be about 37% greater than those b u i l t with convent ional mortar . Drysdale and Hamid (Ref. 37) found that a decrease i n mortar strength of about 70% r e s u l t e d i n a corresponding decrease i n prism strength of l e s s than 10%. Sah l in (Ref. 4) quotes an i n v e s t i g a t i o n by Nylander in which sand f i l l e d j o i n t s produced a masonry with a strength of about 60% of masonry with medium strength mortar. The in f luence o f the j o i n t th ickness r e l a t i v e to the height of the masonry un i t has a lso been recognized as a s i g n i f i c a n t parameter of the prism s t rength . Since the mortar i s usua l ly the weakest par t of the assemblage, the h ighest strengths are obtained with t h i n bed j o i n t s . The r e s u l t s obtained i n Ref.. 37 showed that i n c r e a s i n g the j o i n t th ickness from 3/8 to 3/4 i n . ( ~10 mm to ~20 mm) resu l ted in the prism strength decreasing by 16% fo r ungrouted masonry (but only 3% fo r grouted masonry). Sah l in (Ref. 4) quotes some i n v e s t i g a t o r s p o i n t i n g out that i f the j o i n t s are t h i n , mortar s t rength has l i t t l e in f luence on the s t rength of the masonry. Masonry wal ls are u s u a l l y l a i d with mortar along the face s h e l l but not along the webs ( f a c e - s h e l l bedding) . Maurenbrecher (Ref. 38) pointed out the need for prisms to r e f l e c t t h i s p r a c t i c e because prisms with face-s h e l l bedding f a i l at an apparent lower s t ress than do those with f u l l bedding. He recommended the use o f the bedded area ins tead of the net 55. area of the blocks in order to obtain the cor rec t s t r e s s e s . If a two b lock pr ism i s tes ted i n compression between p l a t e s or capp-ing mater ia l of higher e l a s t i c modulus than the specimen, there i s an end r e s t r a i n t and the t e n s i l e s t resses induced i n the b locks are reduced. The f a i l u r e load i s increased and f a i l u r e usua l ly takes p lace i n a shear mode s i m i l a r to that observed i n the i n d i v i d u a l block t e s t but not observed in wal ls or prisms composed of more courses . I t can be expected that by i n c r e a s i n g the number of c o u r s e s , the blocks i n the center of the prism would be f ree from the end r e s t r a i n t and be more representa t ive of the cond i t ions i n an a c t u a l w a l l . However, experimental data does not seem to conf irm t h i s expecta t ion . B . F . Boult (Ref. 39) determined that the reduc-t i o n i n s t rength with height appears to be i n s i g n i f i c a n t f o r 3 to 12 course grouted pr isms. Maurenbrecher (Ref. 38) found that the reduct ion i n strength f o r prisms b u i l t with a h e i g h t - t o - t h i c k n e s s r a t i o higher than two (up to h / t = 6) was very sma l l . He suggested that c o r r e c t i o n f a c t o r s app l ied to hollow concrete b lock prisms with h / t > 2 should not be l a r g e r than one (the fac tor fo r h / t = 2 ) . Prisms are u s u a l l y capped to ensure a more uniform load d i s t r i b u t i o n . The masonry standards (eg. ASTM-C447) spec i fy a s u l f u r or a dental p l a s t e r capping. Maurenbrecher (Ref. 38) compared the r e s u l t s f o r prisms tes ted with f i b r e board capping to those obtained for prisms tes ted with p l a s t e r capping. The mean strengths were s i m i l a r , with f i b r e board capping g i v i n g s l i g h t l y lower s t rengths . S imi la r r e s u l t s were obtained by Roberts (Ref. 31) when comparing f i b r e board and mortar caps . Drysdale and Hamid (Ref. 37) inves t iga ted the in f luence of grout ing the pr isms. They determined that the grout , which occupied approximately 56. 40% of the gross a re , d id not contr ibute p r o p o r t i o n a l l y to the prism s t rength . Large increases i n the grout strength r e s u l t e d i n only r e l a t i v e l y small increases in prism s t rength . The t e s t r e s u l t s ind ica ted that at f a i l u r e the b lock i t s e l f was s t ressed to about 80% of i t s compres-s ive strength for ungrouted prisms and to about 60% i n grouted pr isms. The authors suggested that matching the deformational c h a r a c t e r i s t i c s of the grout and block may be more e f f i c i e n t than increas ing the grout strength as proposed by some codes ( e . g . : ACI-531). Drysdale and Hamid (Ref. 37) a lso studied the in f luence of j o i n t reinforcement i n grouted and ungrouted specimens. Use of No. 9 gage wire j o i n t reinforcement r e s u l t e d in an increase of 2% and 5% for the ungrouted and grouted prisms r e s p e c t i v e l y . On the other hand, Ha tz in iko las et a l . (Ref. 13 and 40) reported a 19% decrease i n the compressive strength of two-block prisms with j o i n t re inforcement . Workmanship has a b i g in f luence on the compressive st rength of masonry. The detr imental e f f e c t of poor workmanship i s due to improper f i l l i n g and t o o l i n g of the j o i n t s . Other f a c t o r s i n f l u e n c i n g the strength of the masonry are : cor ing (Refs. 4, 39, 37) , age (Refs. 4, 38) , loading rate (Ref. 38) , and i n i t a l r a t i o of absorpt ion (Ref. 4 ) . 4 .4.2 Un i t Test Method Rather than conduct expensive prism t e s t s to determine the masonry design s t rength , a value of f may be determined using the u n i t t e s t m method. In t h i s case , f* i s obtained from a t a b l e , as a func t ion of the m masonry u n i t s t rength and the type of mortar , and genera l ly produces a 57. conservat ive estimate of f * . I t assumes that mortar bed j o i n t s w i l l be m 3/8 i n . (~10 mm) ( ± 1/8 i n . [±3 mm]) in th ickness . The masonry un i ts and other mater ia ls must be tes ted i n accordance with a p p l i c a b l e CSA Standards [or ASTM when using other Codes: ACI (Ref. 21), UBC (Ref. 23) ] . The f m value obtained from those tab les i s not a f fec ted by whether the block i s hollow or grouted, but the code requ i res fo r the l a s t case that the grout be at l e a s t as strong as the b l o c k s . (CSA-CAN3-S304, C I . 4 . 3 . 3 . 7 ) . 4 .4.3 Pr ism Test at the U n i v e r s i t y of B r i t i s h Columbia Sixteen two-block prisms were b u i l t at U . B . C . for the f i r s t s e r i e s of t e s t s . One prism b u i l t with h a l f b lock u n i t s was grouted, the o t h e r s , b u i l t with s t re tcher or end block u n i t s , were a l l ungrouted. Some of the ungrouted prisms were f u l l y bedded, the r e s t were constructed with f a c e -s h e l l mortar o n l y . The un i ts used for the const ruc t ion of the prisms were se lec ted randomly from the b locks used i n the wal ls specimens, and the prisms were l a i d by the masons who constructed the w a l l s . The prism j o i n t s were too led i d e n t i c a l l y to the wal l j o i n t s . The prisms were l e v e l l e d and plumbed. For ease of l e v e l l i n g each prism was b u i l t on a mortar bed, and p laced i n s i d e a polyethylene bag to preserve the moisture and reproduce the cur ing cond i t ions i n the wal l j o i n t s . The p l a s t i c bags were opened jus t before capping. The prisms were capped on or the day befqre the t e s t day with Hydrostone Gypsum cement. Although care was taken dur ing the handl ing of the pr isms, the bond between the b locks and the mortar was broken i n two of the e a r l i e r specimens. Table 4.IV(a) shows the strengths obta ined . I f the value of f* was to be determined by the u n i t and mortar t e s t s m 58. method using Table 12 i n CSA-CAN3-S304, we would have obta ined: f = 1235 p s i = 8.5 MPa ( for mortar type N, and block compressive m strength of 3520 p s i = 24.3 MPa). which i s much smal ler than the 2000 ps i (13 .8 MPa) obtained exper imenta l ly . In an attempt to inves t iga te the in f luence of a l te rna te methods of capping on the pr ism t e s t r e s u l t s , three exploratory t e s t s were c a r r i e d out using Doha Cona boards, a moderately sof t f i b r e board. In two t e s t s the capping cons is ted of 1/2" Dona Cona board cover ing the e n t i r e contact sur face , and one t e s t used 3/8" Dona Cona on the face s h e l l s o n l y . In a l l three cases a t e n s i l e s p l i t t i n g f a i l u r e mode with v e r t i c a l cracks running through the block webs was observed. Test r e s u l t s shown i n Table 4.VI(b) confirm the observat ion made by Maurenbrecher (Ref. 38) that f u l l capping causes premature crack ing as a r e s u l t of bending i n the web sect ions not bedded i n mortar , although i n our case the th inner capping board may have contr ibuted to the higher s t rength . 4.5 Re in fo rc ing S t e e l 4 .5.1 Re in fo rc ing Bars Bars used to r e i n f o r c e masonry are normally the same as used i n r e i n -forced concre te , and thus must comply with ASTM Standard A615 - Grade 40 or 60. As the reinforcement i s p laced mainly to r e s i s t t e n s i l e fo rces and provide d u c t i l i t y , i t s most important parameters are the y i e l d s t r e s s , the u l t imate strength and i t s e l o n g a t i o n . Genera l ly grade 40 i s recommended for i t s greater d u c t i l i t y but i n circumstances where there are very la rge l o a d s , grade 60 might be used. 59. 4 .5 .1(a) Tests at the U n i v e r s i t y o f B r i t i s h Columbia The reinforcement bars used at U . B . C . cons is ted of #4, and #6 grade 60 b a r s . The t e s t specimens cons is ted of segments of the #4 and #6 bars a c t u a l l y used i n the w a l l s . Table 4.VII presents the r e s u l t s . 4 .5 .2 J o i n t Reinforcement The Canadian Code (CSA-CAN3-S304 i n a r t i c l e s 4 . 6 . 8 . 1 . 2 . and 4 .6 .8 .2 .4) s p e c i f i e s that wire reinforcement i n the mortar j o i n t s may be considered as requi red h o r i z o n t a l s t e e l . J o i n t reinforcement c o n s i s t s of l o n g i t u d i n a l wires jo ined with in termi t tent wires in e i t h e r a ladder or t russ type arrangement. 4 .5.2(a) Tests at the U n i v e r s i t y of B r i t i s h Columbia The j o i n t reinforcement used at U.B.C was galvanized #9 gauge wire ladder type, with the wires bent in to a s l i g h t l y corrugated shape. T e n s i l e t e s t s were performed on a s i n g l e wi re , without any surrounding mortar, and so the corrugat ions would tend to s t ra ighten out more than i n an ac tua l w a l l . Test r e s u l t s are summarized i n Table 4 .VI I I . 60. TABLE 4.1 Compression Strength of Concrete Blocks Sample # Max. Load ( l b s . ) St ress on Gross Area (119 i n 2 ) S t ress on Net Area (63 i n 2 ) 1 272,000 2283 4317 2 257,000 2157 4079 3 220,000. 1847 3492 4 276,500 2321 4381 5 231,000 2107 3984 AV! ERAGE 2143 p s i (14.8 MPa) 4051 (27.9 MPas) STANDARD DEVIATION 187 p s i (1.3 MPa) 352 p s i (2.4 MPa) COMPRESSIVE BLOCK STRENGTH 1865 p s i (12.9 MPa) 3520 p s i (24.3 MPa) Table 4.I I Tension Test of Blocks T e s t . St ress (psi) 1 268 2 152 3 180 4 225 5 209 Average 207 (1.4 MPa) 61. Table 4 . I l l - Par t I Mortar Compressive Strength Test # Age Compressive Strength Mortar Components (Days) (psi) cement : l ime :sand 1 7 days . 1090 1 : 1/2 : 5 2 I t i t 900 3 28 days 1665 I I •1 , 4 I I II 1660 " 5 •1 H 1430 I I II 6 . I I • i 1530 1 : 1/2 : 5 7 •1 i i 1550 1 : 1/2 : 4 1/2 8 •1 n 1515 11 •1 II 9 I I t i 1415 •1 t l II 10 I I II 1020 •1 II n 11 I I i t 995 I I II i t 12 •I • i 1000 1 : 1/2 : 4 1/2 13 11 •t 635 1 : 1/2 : 5 1/2 14 t l i t 650 I I I I I I 15 • • • i 850 i t •1 •1 16 •1 •t 850 i t II t l 17 II t i 890 II II II 18 II t i 895 t i •1 II 19 II i i 900 i t •1 20 •I • i 885 • i I I I t 21 I I II 880 i i II I I 22 • • t i 855 t i I I II 23 II II 910 1 : 1/2 : 5 1/2 Average of 21 t e s t s at 28 days: 1095 p s i (7.6 MPa) Note: Changes i n the sampling procedure were c a r r i e d out from sample onwards. 62. Table 4 . I l l - Par t II Mortar Compressive Strength (Samples Corresponding to Wall Ser ies Not Tested i n Th is Program) Test # Age (Days) Compressive Strength (psi) Mortar cement Components : l ime:sand 24 37 1470 1 : 1/2 : 5 25 II 1740 II •I •1 26 • • 1395 •1 II •1 27 n 1580 II II II 28 it 1410 II II •1 29 II 1655 •1 11 II 30 36 1085 II II •I 31 II 1045 •1 II II 32 1050 II II •I 33 II 1175 II II II 34 II 1205 II •1 II 35 II 1220 1 : 1/2 : 5 Average of l a s t 12 t e s t s : 1335 p s i (9.2 MPa) 63. Table 4.IV - Par t I Mortar Tens i le -Bond Strength Test # Age (Days) T e n s i l e Strength (psi ) Mortar Composition Cement:Lime:Sand 1 45 26 1 : 1/2 : 4 1/2 2 •1 31 II II •1 3 n 20 II II •1 4 n 42 1 : 1/2 : 4 1/2 5 28 61 1 : 1/2 : 5 1/2 6 I I 48 II 11 •1 7 •i 73 II II •1 8 I I 80 •I II II 9 n 96 It II II 10 I I 112 •I •1 II 11 • • 82 II II II 12 i i 71 •1 II II 13 • • 63 II II •I 14 II 48 •I •1 II 15 I I 60 1 : 1/2 : 5 1/2 Average of 15 t e s t s : 61 p s i (0.4 MPa) Table 4.IV - Par t I I Mortar Tens i le -Bond Strength (Samples Corresponding to Wall S e r i e s Not Tested i n Th is Program) Test # Age (Days) T e n s i l e Strength (psi ) Mortar Cement Compression :Lime:Sand 1 37 28 1 : 1/2 : 5 2 •I 63 •1 II II 3 43 39 II •I •1 4 •i 95 •1 II •I 5 I I 25 II II •1 6 • i 65 II II II 7 II 62 II II II 8 • • 59 •1 II II 9 n 100 II •I II 10 II 120 If II II 11 • i 35 II •1 II 12 I I 51 •I II II 13 II 54 II II II 14 I I 46 1 : 1/2 : 5 Average of 14 t e s t s : 60 p s i (0.4 MPa). 64. Table _4 .V - Par t I Grout Compressive Strength Test Age (Days) Compressive Strength (psi) Composition 1 28 2722 cement 19 lbs 2 II 2736 coarse sand ->• 88 l b s 3 •1 2815 3/8" stone + 35 l b s 4 31 3006 Pozz -*• 20 ml 5 II 3072 6 II 3386 7 28 3229 8 II 2625 9 n 2444 10 •i 4285 11 it 3628 12 •i 3821 13 • i 3335 14 •i 3893 Average for 14 t e s t s = 3214 p s i = 22.2 MPa. Slump f o r samples 1 to 3 •»• 7.5" = 190 mm. M •> " 4 to 8 6 " = 150 mm. Note: the propor t ion of sand was o r i g i n a l l y 83 l b s , but was increased to 88 lbs i n order to c o r r e c t fo r b u l k i n g . Table 4.V - Par t II Grout Compressive Strength  (Sample Correponding to a Walls Ser ies Not Tested Yet i n the Program) Test Age Compressive Strength (psi) Composition 15 37 3560 cement + 19 l b s 16 37 3635 Coarse Sand + 83 l b s 17 36 2730 3/8" •>• 35 l b s 18 • i 2575 Pozz •*• 20 ml 19 • • 3940 Average Compressive Strength for Samples 15 to 19 = 3288 p s i - 22.7 MPa. Slump fo r Samples 15 to 19 = 7 1/4" = 185 mm. 65. Table 4.VI(a) Prism Tests Test Age (Days) Prism Strength (psi) Notes 1 36 2302 No bond between mortar and blocks 2 35 2381 3 35 2683 4 35 2349 No Bond between mortar and blocks 5 35 2167 6 II 2294 7 •1 2516 • 8 It 2079 9 It 2603 10 47 2214 f a c e - s h e l l bedded * 11 II 1913 II H •• * 12 46 2008 II ii II * 13 It 2476 14 II 2365 15 tl 2429 16 132 3,776 p s i (was kept i n moisture room) h a l f b lock , grouted f* based on grouted area m grout : 2 1/2 sand, 1 1/2 pea gravel 1 cement Average of 15 ungrouted t e s t s : 2319 p s i = 16 MPa. Th is would produce f ' m = 2000 p s i = 13.8 MPa. The r e s u l t s are based on a net area of 63 i n . *If based on the net bedded area (as recommended i n Ref. 38) of 48 i n 2 we would obta in 2906 p s i instead of 2214 p s i , 2511 instead of 1913 p s i and 2636 instead of 2008 p s i . Table 4.VI(b) Pr ism Tes ts Capped with Dona Cona Boards Test Capping board / th ickness ( in . ) age (days) Compressive strength (psi) (based on block net area 63 i n 2 ) 1 1/2 92 626 2 1/2 II 745 3 3/8 •i 1505* Note: *When only face s h e l l i s taken in to account: strength - 2230 p s i 66. Table 4.VII Tension Tests of Re in fo rc ing Bars Test Tens i l e Stres #4 (psi) T e n s i l e Stress #6 (psi ) Y i e l d Max. Y i e l d Max. 1 66500 106750 2 66500 107200 3 4 67500 105800 70909 112045 5 70114 110000 6 71364 110227 7 68750 107000 8 9 69000 107500 64773 10 6477 10 66591 108409 11 65000 10 6364 The average y i e l d s t r e s s of the eleven specimens was: 67900 p s i (470 MP The average ul t imate strength of the eleven specimens was: 108000 p s i (745 MPa) Note: On average the #6 bars were s l i g h t l y stronger than the #4 bar Table 4.VIII Tension Tests of Wire Reinforcement Specimen Y i e l d Stress O y (psi ) Ult imate Stress ° u l t ( P s i ) 1 58,365 60,699 2 — 78,500 3 49,610 64,200 4 55,450 67,120 5 74,120 75,290 Average 59,400 69,200 67. CHAPTER 5 Test S e r i e s at the U n i v e r s i t y of B r i t i s h Columbia At the U n i v e r s i t y of B r i t i s h Columbia a t ten t ion i s being d i r e c t e d i n i t i a l l y to the out of plane forces ac t ing on non- loadbear ing w a l l s . The t e s t s e r i e s was designed to determine the parameters governing the spacing of p r i n c i p a l reinforcement and the design of d i s t r i b u t i o n s t e e l . The main aims of t h i s study are: (a) To e s t a b l i s h the maximum spacing of main s t e e l without any t ransverse d i s t r i b u t i o n s t e e l fo r wal ls spanning e i t h e r h o r i z o n t a l l y or v e r t i -c a l l y . (b) To determine the e f f e c t s of d i s t r i b u t i o n s t e e l on the spacing of the main s t e e l . (c) To determine the e f f i c i e n c y of j o i n t reinforcement as h o r i z o n t a l d i s t r i b u t i o n s t e e l or as main s t e e l for hor i zon ta l spanning w a l l s . (d) To determine a method of p r e d i c t i n g the a b i l i t y of the masonry to span between the main reinforcement or l a t e r a l suppor ts . The f i r s t ob jec t i ve was to determine an upper l i m i t to the spacing without d i s t r i b u t i o n s t e e l . Thus, specimens having reinforcement s t e e l i n one d i r e c t i o n only were t e s t e d , s t a r t i n g with wal ls spanning v e r t i c a l l y with i n c r e a s i n g spacing between re inforcement . Var ious arrangements of d i s t r i b u t i o n s t e e l i n the form of bond beams or j o i n t reinforcement were then introduced i n order to study t h e i r in f luence i n the behaviour of the w a l l s . Tests were a l s o performed on wal ls spanning h o r i z o n t a l l y with d i f f e r e n t arrangements of main re inforcement , i n c l u d i n g j o i n t r e i n f o r c e -ment, but with no d i s t r i b u t i o n s t e e l . 68. 5.1 Test Set-Up Test specimens c o n s i s t e d of 14 w a l l s , 12 of which were 8 x 8 f t (2.44 x 2.44 m), one 8 inches (0.20 m) higher and one 16 inches(0.40 m) shor te r . A l l were b u i l t of 8 x 8 x 16 i n . (200 x 200 x 400 mm) ungrouted hollow concrete b locks i n running bond c o n s t r u c t i o n . Reinforcement s t e e l was grade 60 (60,000 p s i or 415 MPa nominal y i e l d s t ress) bars i n grouted cores or bond beams and #9 gauge ladder - type j o i n t re inforcement . The j o i n t reinforcement was galvanized deformed wire bent i n a s l i g h t l y corrugated pa t te rn . The wal l specimens were constructed and a i r cured i n a labora tory environment. A l l wal ls were constructed by experienced masons using t e c h -niques t y p i c a l of good workmanship and s u p e r v i s i o n . The mortar j o i n t s on both faces were t o o l e d . The wal ls were tes ted i n the v e r t i c a l p o s i t i o n and loaded by means of an a i r bag placed between the panel and a reac t ion w a l l . In order that they could be moved around the laboratory they were a l l b u i l t on a wooden base beam. In an e f f o r t to prevent a l a t e r a l reac t ion from developing at the base of h o r i z o n t a l l y spanning wal ls they were, with two except ions , supported v e r t i c a l l y on s l i d i n g t e f l o n base pads placed between the wal l and the wooden beam. Hooked f l a t s t e e l reac t ion bars p laced at 8 inches (200 mm) on c e n t r e s , and bear ing on the unloaded face of the p a n e l , provided a simple support r e a c t i o n c o n d i t i o n , except that i n the case of the wal ls spanning v e r t i c a l l y t h e i r se l f -we ight gave some r o t a t i o n a l r e s i s -tance at the bottom. L a t e r a l displacements were measured at 9 po in ts on a 3 x 3 g r i d using taut wires d r i v i n g rotary potent iometers. The l a t e r a l pressure was 69. measured with a pressure t ransducer; and i n most cases two s t r a i n gauges were p laced on every main r e i n f o r c i n g bar . The record ing of the data was done by a mul t i -channel V idar "5200 Ser ies D-DAS" recorder d r i v i n g a paper tape. F igure 5.1 shows the t e s t arrangement. 5.2 Test Ser ies To date the t e s t s have invo lved monotonic q u a s i - s t a t i c loads except for two wal ls subjected to one load r e v e r s a l a f t e r y i e l d i n g i n the f i r s t d i r e c t i o n . E ight wa l ls spanned v e r t i c a l l y and s ix h o r i z o n t a l l y . There were 15 t e s t s as wal l #1 was tes ted again a f t e r shear ing o f f the upper course i n the f i r s t t e s t . Table 5.1 shows the wal l dimensions, r e i n f o r c i n g and boundary c o n d i -t ions along with the f a i l u r e p ressure , and a short d e s c r i p t i o n of the mode of f a i l u r e . In some cases a load d e f l e c t i o n p l o t i s a l s o shown. The s t e e l r a t i o , was def ined as p = A / t L , where A i s the t o t a l s t e e l area i n the s s length L and t i s the wal l t h i c k n e s s . A l l the wal ls with v e r t i c a l main s t e e l have 2#6 bars (19 mm diameter) fo r a s t e e l r a t i o of 0.0011. The wal ls with h o r i z o n t a l main s t e e l inc luded #4 (13 mm diameter) and #6 (19 mm diameter) bars i n bond beams, or j o i n t re inforcement . The h o r i z o n t a l s t e e l r a t i o s va r ied from 0.00025 to 0.0014. A l l the r e i n f o r c i n g bars were placed i n the middle of the w a l l s . 5.3 Test Resul ts 5.3.1 V e r t i c a l Spans Accord ing to the Canadian Code, f o r non- loadbear ing wal ls (CSA-CAN3-S304 - Clause 4 . 6 . 8 . 2 . 1 . ) i n seismic zone 3, a minimum s t e e l area of 0.001 70. A (or a s t e e l r a t i o of 0.001) i s required at a minimum spacing of 16 g inches (400 mm) i f there i s reinforcement i n one d i r e c t i o n o n l y . As a s t a r t i n g point t h i s minimum s t e e l area was provided but the spacing was increased to 4 fee t (1.22 m), corresponding to the maximum spacing s p e c i -f i e d for wal ls having s t e e l p laced i n two d i r e c t i o n s . No t ransverse d i s -t r i b u t i o n s t e e l was i n c l u d e d , as i t was des i red to f i n d the maximum spacing of the bars that w i l l b r i n g about a f a i l u r e of the p l a i n masonry between the b a r s , thus e s t a b l i s h i n g the maximum spacing of the main v e r t i c a l s t e e l without any t ransverse s t e e l . The recorded pressures were not p a r t i c u l a r l y accurate fo r the f i r s t 3 t es ts which were exploratory i n nature, but the r e s u l t s from t e s t [1] and [2] with main v e r t i c a l s t e e l at 48 i n . (1.22 m) showed that f a i l u r e d id not occur in the t ransverse d i r e c t i o n i n the blocks between the v e r t i c a l b a r s . Wall [1] f a i l e d i n shear and bond at the top course at a maximum pres-sure of 310 psf (14.8 kPa) . Af ter removing the upper course , wal l [1] was tes ted fo r a second t ime. The w a l l , which was a l ready cracked a f te r the f i r s t t e s t , f a i l e d again along the top course at a pressure of 365 psf (17.5 kPa) . Wall [2] with the same arrangement as wal l [1], f a i l e d at a pressure of 250 psf (12.0 kPa); i t was a bending f a i l u r e of one c a n t i l e v e r por t ion about the v e r t i c a l re inforcement . The spacing between v e r t i c a l b a r s , was then increased i n stages to a maximum of 72 inches (1.83 m). Wall [4] showed that the masonry was able to span at l e a s t 56 inches (1.42 m) between the main s t e e l at the high load of 285 psf (13.6 kPa) . The f a i l u r e mode was s i m i l a r to that of wa l l [2] , a bending f a i l u r e of one s ide about the v e r t i c a l re inforcement . 7 1 . Wall [5] with a s t e e l spacing of 72 inches (1.83 m) and small edge c a n t i l e v e r f i n a l l y f a i l e d by a bending mechanism i n the masonry between the main r e i n f o r c i n g b a r s , although at a high load of 193 psf (9.2 kPa) . In an exploratory t e s t to study the e f f e c t s of repeated l o a d i n g , an attempt was made to simulate the ac t ion of a cracked wal l by reducing or dest roy ing the bond between b locks and mortar through the a p p l i c a t i o n of a bond breaker . To t h i s end wal l [6] was b u i l t with the mortar faces of the b locks dipped i n Sternson Bond Release, a compound commonly used to prevent bond between l i f t s labs poured on top of one another. The main s t e e l con-s i s t e d on 2#6 bars arranged s i m i l a r l y to wal l [5]. A t ransverse #4 bar was placed i n the top course to provide some containment c a p a c i t y . I t was hoped that the r e s u l t of t h i s t e s t would give a lower bound on the beha-v iour under c y c l i c l o a d i n g . However, the f a i l u r e mechanism and load were e s s e n t i a l l y the same as f o r wal l [5] (210 l b / f t 2 = 10.0 kPa) . It i s not known whether the bond breaker f a i l e d to perform or whether cracks have l i t t l e e f f e c t on the load r e s i s t a n c e . Wal ls [5] and [6] showed that the masonry would r e s i s t a f a i r l y high l o a d , about 200 psf (9.6 kPa) over a span of 72 inches (1.83 m) between l i n e s of main re inforcement . To inves t iga te how much the span could be inc reased , and to learn how h o r i z o n t a l d i s t r i b u t i o n reinforcement would a f f e c t the f a i l u r e mode, severa l wal ls were b u i l t with increase d is tance between the v e r t i c a l main s t e e l p lus h o r i z o n t a l d i s t r i b u t i o n s t e e l . The amount of d i s t r i b u t i o n s t e e l was designed so that the p red ic ted f a i l u r e mode would be a mechanism i n the b locks or y i e l d i n g of the d i s t r i b u t i o n s t e e l , but not a f a i l u r e of the main s t e e l . D i s t r i b u t i o n s t e e l i n the form of j o i n t reinforcement or bond beams was provided with main s t e e l at 72 72. inches (1.83 m) i n wal ls [7] and [8] and 88 inches (2.24 m) i n wal l [14]. In wal l [7] , the j o i n t reinforcement was p laced i n every course i n order to obtain the same reinforcement r a t i o as #4 bars at 48 inches . (1 .22 m), which was the hor i zon ta l reinforcement for wal l [8]. A #4 bar was p laced at the top to prevent a premature shear or bond f a i l u r e i n the top course . Wall [7] f a i l e d by a bending mechanism i n the b locks between main b a r s . Wall [8] f a i l e d by bond on the hor i zon ta l #4 bar at mid-he ight . In both cases the primary bending capaci ty was a lso e s s e n t i a l l y f u l l y deve l -oped. Wall [7] with the j o i n t reinforcement performed bet te r in the sense that there was l e s s s ign of d i s t r e s s before f a i l u r e . A f t e r t h i s good per -formance by wal l [7] , i t was decided to b u i l d wal l [14] with the main ver -t i c a l s t e e l at 88 inches (2.24 m) spacing and d i s t r i b u t i o n s t e e l c o n s i s t i n g of j o i n t reinforcement every second course . This wal l f a i l e d i n t ransverse bending between the main b a r s . The f a i l u r e load f o r wa l ls [7] and [8] was more or l e s s the same and about 1.75 times that of wal l [5] which d id not have the d i s t r i b u t i o n s t e e l . Wall [14] had about the same f a i l u r e load as wal l [5]. F igure 5.2 shows the d e f l e c t i o n of the centre point with respect to the four corners f o r wal l [14]. 5.3.2 H o r i z o n t a l Spans The h o r i z o n t a l l y spanning wal ls inc luded main s t e e l spacing of 48 inches (1.22 m) (wal ls [3] and [9]) and 72 inches (1.83 m) (wal l [10]) , and wal ls with only j o i n t reinforcement as main s t e e l (wal ls [11], [12], [13]). No v e r t i c a l d i s t r i b u t i o n s t e e l was used in these w a l l s . In wal ls [12] and [13] j o i n t reinforcement was p laced i n every course ( i e . a t 8 inches = 200 73. mm), whereas in wal l [11] i t was p laced i n every second course (at 16 i n c h e s ) . In order to see whether the t e f l o n pads changed the mode o f f a i l -ure or f a i l u r e load from that of a wal l r e s t i n g on the wooden beam (or the f l o o r ) , they were omitted at the base of wal l [13]. In a l l these cases (walls [9], [10], [11], [12], [13]) the f a i l u r e load was very near ly the same: 125 to 130 p s f (6.2 kPa) . Wall [3] f a i l e d at a s l i g h t l y higher load (160 psf = 7.7 kPa) . In a l l c a s e s , there was no c rack ing and very l i t t l e deformations up to the f a i l u r e l o a d , whereafter the load capac i ty dropped o f f with inc reas ing d e f l e c t i o n s . A l l the wal ls showed considerab le d u c t i l i t y . F igures 5.3 to 5.7 show a s e r i e s of p l o t s of the load against the d e f l e c t i o n of the centre po in t of the wal ls r e l a t i v e to t h e i r four c o r n e r s . Wall [3] su f fe red a bond f a i l u r e of the top bar a l lowing a c a n t i l e v e r type of f a i l u r e of the top p o r t i o n . Wall [9] f a i l e d in shear along the bottom of the upper course and then f a i l e d i n a bending mechanism i n the b l o c k s . Wall [10] f a i l e d i n a bending mechanism of the b l o c k s . The d e f l e c t i o n s at the centre po in t with respect to both a v e r t i c a l l i n e and a hor i zon ta l l i n e showed c lose agreement which seems to conf irm a two-way a c t i o n (see F igure 5 .8 ) . Wal ls [11], [12] and [13] appeared to f a i l i n primary bending. The d e f l e c t i o n of wa l ls [11] and [12] confirmed the observat ion of one way bending ( e . g . : see F i g . 5 .9 ) . Wall [13] exh ib i ted more two way bending than wal l [12] i n that there was l e s s damage to the lower par t of the wal l However t h i s d i d not appear to a f f e c t the load as the load d e f l e c t i o n p l o t for the two wal ls were near ly i d e n t i c a l out to about-0.4 inches (10 mm) d e f l e c t i o n where wal l [12] was reversed . 74. Walls [11] and [12] a f te r being tested to f a i l u r e i n one d i r e c t i o n were reversed and loaded from the other s i d e . F igure 5.10 shows that post -crack ing strength was only s l i g h t l y reduced in the reverse c y c l e . S imi la r r e s u l t s were obtained by Scr ivener (Ref. 8 ) . 5.4 A n a l y s i s of Results  5.4.1 Bending 5.4.1(a) F i r s t Cracking Of the wal ls spanning h o r i z o n t a l l y , specimens [11], [12] and [13] conta in ing only j o i n t re inforcement , reached a load of 120 - 130 psf (~ 6.2 kPa) , when cracks formed in the v e r t i c a l and connect ing h o r i z o n t a l mortar j o i n t s . In each case the crack was conf ined to the mortar j o i n t s and d i d not pass through any b l o c k s . A f te r f i r s t c rack ing the load e i ther remained near ly uniform or dropped o f f i n d i c a t i n g an under - re in forced c o n d i t i o n . Wal ls [9] and [10] with bond beams, cracked at approximately the same load as wal ls [11], [12], [13]. In wal l [10] f i r s t c rack ing was immediate-l y fo l lowed by the formation of a mechanism between r e i n f o r c i n g b a r s , and l o s s of load c a p a c i t y . Wall [9] had more c l o s e l y spaced s t e e l , but f i r s t c rack ing was fol lowed by a shear f a i l u r e along the top course fol lowed by a bending f a i l u r e i n the top h a l f of the pannel . In a l l these w a l l s , then, f i r s t c rack ing co inc ided with maximum l o a d , and the c rack ing moment about a 75. v e r t i c a l axis was about 1,000 l b f t / f t (4,450 Nm/m).* In the wal ls spanning v e r t i c a l l y , i t was on ly p o s s i b l e to determine the c rack ing load fo r wal l [14]. I t was 55 ps f (2.6 kPa) , fo r a c rack ing moment o f 440 l b f t / f t (~ 2,000 Nm/m). Th is moment corresponds to a t e n s i l e s t r e s s of about 63 p s i (0.4 MPa) between block and mortar, which compares c l o s e l y with the measured value o f 61 p s i (see Chapter 4 ) . In the case of h o r i z o n t a l spans the crack ing corresponds to a t e n s i l e s t r e s s of 140 to 150 p s i (1.05 MPa) which i s greater than twice the bond t e n s i l e strength of the mortar-block assemblage. Th is d i f f e r e n c e i s a t t r i b u t e d to the running bond const ruc t ion which, i f the crack i s not a s t r a i g h t v e r t i c a l crack through the mortar and b l o c k s , forces the crack through the mortar i n a combination of tens ion and shear f a i l u r e . Other f a c t o r s such as shear f r i c t i o n between b l o c k s , se l f -we ight of the w a l l s , strength of the b l o c k s , and r e s t r a i n t at the wal l base a l l contr ibute to the moment capac i ty of the w a l l . 5.4.1(b) Capaci ty of P r i n c i p a l Reinforcement The measured average s t r a i n s i n the bars were compared to the s t r a i n s c a l c u l a t e d on the b a s i s of a l i n e a r s t r e s s - s t r a i n d i s t r i b u t i o n of a cracked s e c t i o n . F igures 5.11 to 5.15 show the p l o t s of s t r a i n s v s . load obtained fo r wal ls with main v e r t i c a l s t e e l . In a l l cases the p red ic ted t h e o r e t i c a l s t r a i n values were very c lose to the experimental v a l u e s . Based on the measured strength of the r e i n f o r c i n g s t e e l the v e r t i c a l l y spanning wal ls *In the fo l lowing d i s c u s s i o n , moments w i l l be def ined i n the vector sense. Thus, fo r example, a moment about a v e r t i c a l a x i s , when the d i r e c t i o n of the spanning i s h o r i z o n t a l , w i l l be r e f e r r e d as to v e r t i c a l moment. 76. could r e s i s t a load of 285 psf (13.6 kPa) at y i e l d (f = 6 8 ks i ) and 440 Y psf (21.1 kPa) at u l t imate (f =108 k s i ) . u F igures 5.11 and 5.12 show the comparison p l o t s fo r wa l ls [4] and [6] r e s p e c t i v e l y . In both cases the maximum load d i d not reach the t h e o r e t i c a l y i e l d value and t h i s i s r e f l e c t e d i n the p lo ts . . I t can be seen that the t h e o r e t i c a l s t r a i n values are genera l ly higher than the measured s t r a i n v a l u e s . Wal ls [7] and [8] (F igure 5.13 and 5.14) exceeded the t h e o r e t i c a l y i e l d l o a d . In both cases the p red ic ted y i e l d load agreed c l o s e l y with the load at which the measured average s t r a i n s reached the y i e l d s t r a i n . A f te r y i e l d , the bars showed the t y p i c a l p l a s t i c deformation of s t e e l . The unloading branch of the bar i n wai l [8] has not been p l o t t e d because the recorded s t r a i n s showed an e r r a t i c behaviour , probably due to damage caused to one of i t s s t r a i n gages a f t e r f a i l u r e in the bond beam. Figure 5.15 shows the p l o t for wa l l [14]. In t h i s case , up to a load of about 140 psf (6.7 kPa) , the measured s t r a i n s are smal ler than the pre -d i c t e d va lues . For higher loads the opposite holds i n a few cases but then the measured values remain wi th in 3% of the t h e o r e t i c a l v a l u e s . Although t h i s bar d id not reach the y i e l d s t ress some r e s i d u a l s t r a i n s are observed i n the unloading branch. They are probably due to a r e l a t i v e movement of the blocks not a l lowing a f u l l recovery of the bar . For wa l ls [7] , [8] and [14] only the average s t r a i n o f one bar was p l o t t e d because the values obtained for the second bar were considered d e f e c t i v e . In general i t can be observed that fo r the wal ls tes ted with v e r t i c a l 77. main s t e e l , the measured s t r a i n values fo l low the same trend as the theore -t i c a l v a l u e s . The l a t t e r are bigger than the measured ones but the gap c loses when approaching the y i e l d l o a d , which was very wel l p red ic ted for wal ls [7] and [8] . Using ul t imate strength theory , Scr ivener (Ref. 8) obtained «s imi la r r e s u l t s for t e s t s on r e i n f o r c e d b r i c k w a l l s . He assumed the s t r e s s - s t r a i n curve f o r the b r i c k to be the same as f o r concrete and used the Whitney equation for r e i n f o r c e d concrete . That there i s not much d i f f e r e n c e between the r e s u l t s obtained using e i ther approach i s due to the l i g h t reinforcement which causes a small compression zone i n the masonry. In an attempt to analyze the behaviour of the h o r i z o n t a l l y spanning wal ls a s i m i l a r comparison p l o t was drawn fo r wal ls [9] and [10] (F igures 5.16 and 5.17 r e s p e c t i v e l y ) . The general behaviour of the r e i n f o r c i n g bars corresponds to the observed behaviour i n the wal ls e x h i b i t i n g small d i s -placement u n t i l the c rack ing load was reached; t h e r e a f t e r , la rge deforma-t i o n s . In these cases the experimental po in ts have l i t t l e correspondence to the t h e o r e t i c a l values obtained assuming a constant curvature across any v e r t i c a l s e c t i o n . I t i s apparent that the d i s t r i b u t i o n of moments across the height of the s e c t i o n i s not uni form. For example, at maximum load the midspan moment of wal l [9] (Figure 5 .16) , i f proport ioned on the b a s i s of s t e e l a reas , would requ i re a moment of 53,650 l b . i n (6,060 Nm) to be r e s i s t e d by the #6 bar and 26,825 l b . i n (3,030 Nm) by each of the #4 b a r s . If we change our assumptions and assume that the wal l behaves as a cont inu -ous s lab i n the v e r t i c a l d i r e c t i o n with the bond beams as supports , the moments i n each bond beam would be d i s t r i b u t e d i n the same propor t ion as the reac t ions on the suppor ts . Thus, the #6 bar would r e s i s t a moment of 67,100 l b . i n (7,580 Nm) and each one of the #4 bars a moment of 20,120 78. l b . i n (2,275 Nm). According to the measured s t r a i n s the moment d i s t i b u t i o n i s as fo l lows: . 13,680 l b . i n (1,545 Nm) by the upper #4 b a r , 28,950 l b . i n (3,270 Nm) by the lower b a r , and 48,760 l b i n (5,510 Nm) by the #6 ba r . Th is moment d i s t r i b u t i o n : (a) does not add up to the t o t a l app l ied moment and (b) con t rad ic ts the d i s t r i b u t i o n based on the above two assumptions and the more l i k e l y moment d i s t r i b u t i o n i n which, the upper #4 bar would have higher s t r a i n s than the lower bars because of i n c r e a s i n g moment r e s i s t a n c e to h o r i z o n t a l bending i n the masonry from top to bottom due to the s e l f -weight of the w a l l , and base r e s i s t a n c e . Inspect ion of the wal l displacements and bar s t r a i n s before f a i l u r e , and the mechanism of f a i l u r e , helps to exp la in the d iscrepancy. Before f a i l u r e , the displacements and s t r a i n s measured were very small with higher values being r e g i s t e r e d at the top. At f a i l u r e (see Table 5.1) the upper course ( i . e . the upper bond beam) separated from the r e s t of the wal l by a hor i zon ta l crack running along the en t i re wal l l eng th . Th is could exp la in the low s t r a i n r e g i s t e r e d i n the upper #4 bar . In the top par t of the wal l between the bond beams, a y i e l d l i n e type of mechanism developed which could have helped to reduce the par t of the moment r e s i s t e d by the middle bond beam. The one way bending mechanism observed in the lower par t of the wal l i s perhaps the cause of the high s t r a i n s r e g i s t e r e d i n the bottom #4 b a r , because i n t h i s type of mechanism the res is tance w i l l be concentrated at the bond beams. In t r y i n g to exp la in why such a high s t r a i n was measured i n the bottom bar , we cannot d iscount the p o s s i b i l i t y that the bar was not centered. We can see how the rather complicated mechanism makes i t exceedingly d i f f i c u l t to estimate the moment d i s t r i b u t i o n along a v e r t i c a l s e c t i o n . 79. Based on the measured strength of the wi re , the j o i n t reinforcement should susta in a y i e l d load of 41 psf (~2.0 kPa) and an u l t imate load of about 48 psf (2.3 kPa) for p = 0.00025 as i n wal l [11], and double t h i s i n wal ls [12] and [13] with p = 0.0005. The load c a r r i e d by wal ls [11], [12] and [13] a f t e r crack ing cons iderab ly exceeded the capac i ty of the j o i n t reinforcement and po in ts out the importance of masonry mechanisms i n r e s i s t i n g the post c rack ing l o a d s . 5 .4 .1(c) Bending Resistance of Blocks Between Reinforcement In order to set l i m i t s on the spacing of the p r i n c i p a l re inforcement , and to be able to design the d i s t r i b u t i o n s t e e l , one must know the beha-v iour of the masonry between the main r e i n f o r c i n g b a r s . Considerable e f f o r t s have been made to f i n d a theory su i tab le for t h i s purpose. Where f a i l u r e d id occur i n t h i s mode, a y i e l d l i n e pat tern was observed to form i n the panels i n s i d e a perimeter def ined by the main s t e e l and the supports (see Table 5 .1 ) . Y i e l d l i n e analyses were c a r r i e d out on those wal ls that showed a wel l def ined bending mechanism. Y i e l d l i n e theory assumes that the bending moment at a po in t along a l i n e reaches a y i e l d value and remains constant while other par ts of the l i n e reach that va lue ; thus , a y i e l d l i n e pat tern develops with constant moments along each l i n e . Th is type of ana lys is impl ies r e l a t i v e l y large d e f l e c t i o n s and d u c t i l e beha-v i o u r , which was observed i n our w a l l s . Assuming that the y i e l d l i n e s fo l low the major crack pat terns as dimensioned i n Table 5. I I , the r e l a t i o n between moment c a p a c i t i e s and l a t e r a l load i s given i n column 5 of Table 5.I I . Let M , = v e r t i c a l moment capac i ty of the b locks vb 80. M = hor i zon ta l moment capac i ty of the blocks H M = v e r t i c a l moment capac i ty o f j o i n t reinforcement every 2nd vr course . M = M + M = t o t a l v e r t i c a l moment c a p a c i t y . V vr vb In ex tens ive ly cracked wal ls the masonry can have l i t t l e res is tance to h o r i z o n t a l moments, only the dead load of the wal l or the in -p lane forces imposed by the v e r t i c a l reinforcement can provide some r e s i s t a n c e . With respect to the v e r t i c a l moment capac i ty i n wal ls with running bond c o n s t r u c t i o n , where the cracks fo l low the mortar l i n e s , curvature must cause r e l a t i v e r o t a t i o n between the i n t e r l o c k i n g b l o c k s . The consequent shear f r i c t i o n could be expected to provide r e s i s t a n c e to v e r t i c a l moments greater than the h o r i z o n t a l moments. The magnitude would be a funct ion of the v e r t i c a l forces a r i s i n g from the dead weight or the clamping force caused by the tension in the v e r t i c a l s t e e l . Previous works (Refs. 2, 3, 4 and 7) reported a modulus of rupture in h o r i z o n t a l bending ( i . e . v e r t i c a l moments) that was 3 to 6 times higher than the modulus of rupture i n v e r t i c a l bending ( i . e . h o r i z o n t a l moments). The observat ions made in our s e r i e s of t e s t tended to confirm the above; the c rack ing load fo r wal l [14] was 55 psf (2.6 kPa) whereas wal l [9] to [13] spanning h o r i z o n t a l l y had a c rack ing load of 120 to 130 psf (~6.2 kPa) . I t i s important to note that whereas a l l the above mentioned r e f e r -ences, and our own r e s u l t s , deal with c rack ing moments (or modulus of rup-t u r e ) , the y i e l d l i n e a n a l y s i s deals with the u l t imate moment c a p a c i t y . W< w i l l assume that the same r e l a t i o n s h i p a p p l i e s fo r the u l t imate moment capac i ty r a t i o , i . e . : M , / M = 3 to 6. 81. The data presented i n Table 5.II suggest that M i s cons iderab ly vb greater in wal ls with v e r t i c a l main s t e e l than i t i s in those with h o r i z o n -t a l reinforcement o n l y . Th is i s cons is ten t with the f a c t that there i s an extra v e r t i c a l compressive force on the blocks a r i s i n g from the tension of the r e i n f o r c i n g b a r s . If i t i s taken that M = 460 l b s . f t / f t (2.07 kNm/m) in wal ls with v e r t i c a l vb reinforcement M = 190 l b s . f t / f t ( 0 . 8 6 kNm/m) i n wal ls without v e r t i c a l r e i n -vb forcement M , = 3 M vb H M = 333 l b s . f t / f t fo r p = 0.00025 vr then the l a t e r a l loads that would be sustained by the assumed mechanisms are l i s t e d i n column 7 of Table 5. I I , and the discrepancy between these values and the observed loads are given i n column 8. Except fo r wal ls [7] and [14] the agreement i s good. Wall [7] may not have been a true t e s t of the y i e l d l i n e theory as the f a i l u r e load a lso exceeded the y i e l d capac i ty of the main re inforcement . In wal l [14] there was a shear f a i l u r e below the top course; i f the estimated force i n t h i s j o i n t i s inc luded i n the a n a l y s i s , by assuming that at l e a s t 2/3 of the weight of the upper course bears on the wal l with a f r i c t i o n c o e f f i c i e n t of 1.0, the e r ro r i s reduced to 13%. If we change our i n i t i a l assumption and assume M , = 6M the moment vb H c a p a c i t i e s that give reasonable agreement with the experimental r e s u l t s are : 82. M = 510 l b . f t / f t ( 2 . 2 9 Nm/m) i n wal ls with v e r t i c a l reinforcement vb M , = 235 l b . f t / f t ( 1 . 0 6 Nm/m) in wal ls without v e r t i c a l reinforcement vb M = 333 l b . f t / f t ( 1 . 5 0 Nm/m). vr I t i s i n t e r e s t i n g to note that the estimated moment c a p a c i t i e s do not d i f f e r much with a change i n the M , / M r a t i o . (See Table 5.II column 9 and * vb H 10). The quest ion a r i s e s of what f r i c t i o n c o e f f i c i e n t (u) would be requi red to achieve such moment r a t i o s . I f we assume that because of simultaneous h o r i z o n t a l bending the contact area between i n t e r l o c k i n g blocks w i l l be reduced to the face s h e l l , then the shear area w i l l be as shown i n F igure 5.18. Assuming a p l a s t i c y i e l d shear s t ress d i s t r i b u t i o n (Ref. 4) as shown in F igure 5.19 the maximum moment c a r r i e d on such an area i s : fcl fcl fcl fcl M = T - i < £ - j t - ) = u o - i ( £ - — ) (1) where T i s the shear s t r e s s , o i s the v e r t i c a l compressive s t r e s s , and t-^  and H are def ined i n Figure 5.19. The moment capaci ty per un i t height would be given by: M fcl fcl \ b = b - * ° ^ < * - - r ) (2) where b i s the block he ight . The moment capac i ty i n v e r t i c a l bending per un i t length i s given by (see Figure 5.20): M = at, d 1 (3) H 1 2b 1 M b where d 1 i s the lever arm. Thus, y = . . —— (4) V r ' H 83, In our case -t = 1.5 i n . I =8 i n . d 1 = 3.5 i n . b = 8 i n . Therefore the c o e f f i c i e n t of f r i c t i o n u requi red would be: for M = 3 M U = 14.9 and for M = 6 M ' u = 2 9 . 9 vb H vb H These values are very large and obviously not r e a l i s t i c . If we e l a -borate a b i t more on our assumption, we can argue that because of h o r i z o n t a l moments, the contact area i s e s s e n t i a l l y reduced to a l i n e , and the shear s t resses ac t on ly i n a d i r e c t i o n perpendicular to the motion (Figure 5 .21(a) ) . The moment c a p a c i t i e s in t h i s case w i l l be given as : i 1 (5) (6) M = qd J H i2 M v b - P q to where q i s the v e r t i c a l compressive load per u n i t length a r i s i n g from v e r t i c a l bending. 4d!b M v b <7) Therefore U = % 2 M H M vb and thus for —— = 3 , U = 5 . 3 H M vb and - — = 6 , V = 10.5 M H These are a lso very high v a l u e s . We can a l s o argue that because of the higher s t i f f n e s s of the block around i t s webs, res is tance to v e r t i c a l bending (M ) w i l l be concentrated H around those areas . Thus, most of the v e r t i c a l load and the r e s u l t i n g 84. shear forces would be concentrated near the webs as ind ica ted i n F igure 5 .21(b) . Thus: M u = 2f 2- (8) H v 1 M V i = f u ^ = p f 7; = y (9) vb H b v b 2bd l M 2bd A vb . . . . Therefore: P = - — (10) £ 2 M H where f i s the v e r t i c a l compressive force a r i s i n g from v e r t i c a l bending as i n d i c a t e d i n F igure 5 .21(b) . M For ~ - = 3 , v = 4.7 M H M vb and f o r = 6 , V = 9.4 M H and again the values obtained are h i g h . From the above d i s c u s s i o n , i t would seem that a M /M r a t i o of 6 vb H would not be a r e a l i s t i c v a l u e , and that even a r a t i o of M , /M T T = 3 i s vb H h igh . The r e s u l t i n g s t r e s s d i s t r i b u t i o n would probably be a combination of the l a s t two cases d i s c u s s e d , i . e . , a contact area reduced to a l i n e and v e r t i c a l and shear load concentrat ion around the webs of the s l i d i n g b l o c k s . We can expect an increase i n the v e r t i c a l compression i n the areas where the blocks s l i d e over one another, as the shear displacement would requ i re d i l a t i o n i n the mortar or at the mortar block i n t e r f a c e . (See F i g . 5 .21(c ) ) . This would produce a higher M ^/My r a t i o for a given c o e f f i c i e n t of f r i c t i o n . I t i s evident that we need a bet ter assessment of the s t r e s s d i s t r i b u t i o n across the wal l s e c t i o n i n order to estimate a value fo r the M , /M r a t i o , vb H 85. On the other hand, the average se l f -we ight of the wal ls at midheight i s about 200 l b / f t (2.92 kN/m) which would provide a h o r i z o n t a l r e s i s t i n g moment capac i ty of approximately 60 l b . f t / f t (0.27 kNm/m). Fol lowing the assumption that M = 3 M t h i s leads to M = 180 l b . f t / f t (0.81 kNm/m) vb H vb which i s c lose to the value used f o r c a l c u l a t i o n s i n Table 5. I I . For those wal ls with v e r t i c a l main s t e e l , the assumed moment capac i ty M = 460 vb l b . f t / f t (2.07 kNm/m) would imply, us ing the same argument as i n the previous paragraph, an average v e r t i c a l force in the h o r i z o n t a l j o i n t s of 550 l b / f t (8.03 kN/m), corresponding to an e f f e c t i v e s t ress i n the main s t e e l of only 3000 p s i (20 MPa), which i s not unreasonable. 5.5 Shear and Bond ( V e r t i c a l Spans) A. Hamid et a l . (Ref. 41) found that the shear strength of ungrouted masonry was not s t rongly r e l a t e d t o . e i t h e r the compressive strength of the mortar or the compressive strength of the masonry assemblage, but rather to the p h y s i c a l p roper t i es of both mortar and blocks such as surface roughness and i n i t i a l ra te of absorp t ion . Grouting was shown to s i g n i f c a n t l y increase the shear strength of masonry j o i n t s and the authors concluded that along with the l e v e l of the normal compressive s t r e s s , the shear strength of the grouted cores i s the dominant f a c t o r i n the shear strength of grouted concrete masonry. Using a regress ion a n a l y s i s the authors pred ic ted the shear strength fo r concrete masonry as f o l l o w s : 86. T = 76 + 1.07 o*n [psi] for ungrouted masonry (net area) T = 114 + 1.08 a [psi] fo r grouted masonry with weak ag n grout (gross area) I (11) T = 156 + 1 . 5 4 o [psi] for grouted masonry with strong ag n grout (gross area) where o = normal compressive s t ress ' J n —i In the U n i v e r s i t y of B . C . s e r i e s of t e s t s , Wall [1] f a i l e d twice i n what seemed to be a combination of bond and shear f a i l u r e along the top course . In order to have an idea of the shear s t resses susta ined by t h i s w a l l , an e f f e c t i v e shear area as shown i n F igure 5.22 was assumed. Table 5 . I l l shows the shear s t r e s s values obtained f o r wa l l [1] . To give an idea of the s t r e s s l e v e l ach ieved, the values were compared to an ul t imate shear s t r e s s of x = 3.5 /F (ps i ) as def ined i n Ref. 42 f o r the ul t imate shear u c s t r e s s requi red fo r the formation of tens ion cracks i n concrete beams i n the region of high shear and low bending moments. In our case f^ i s assumed to equal the compressive strength of the grout . Comparison was a l s o made with the minimum strength for strong grout of 156 p s i (1.08 MPa) as pred ic ted by Hamid et a l . (Ref. 41) . The anchorage behaviour of grouted masonry depends on the b o n d - s l i p r e l a t i o n s h i p of the bars and grout , and on the res is tance of the grout to s p l i t t i n g . I t a l s o depends on the c o n f i n i n g e f f e c t s of the blocks i n pre -vent ing s p l i t t i n g of the grout core . Cheema and K l inger (Ref. 42), recom-mend caut ion when apply ing concrete anchorage data to grouted masonry, p a r t i c u l a r l y when l a r g e - s i z e bars are used. Despite t h i s , the bond s t r e s s was c a l c u l a t e d using the methods of r e i n f o r c e d concrete (Refs . 26, 27, 42: u = V / E j d ) . Reference 42 gives the ul t imate bond s t r e s s as u = l l / f V d o cr c b J 87. ( p s i ) , where again f i s taken as the compressive strength of the grout . c Table 5.IV shows the bond s t resses c a l c u l a t e d for wal l [1] for the sec t ion between the top two courses and t h e i r comparison to the ul t imate s t r e s s as def ined above. I f we assume that the u l t imate values used in Tables 5 . I l l and 5.IV are c o r r e c t , the fac t that ne i ther the shear nor the bond s t resses reached these values seems to conf i rm the observed combined mode of f a i l u r e . Unfortunately we do not have enough data to develop an i n t e r a c t i o n equation to analyze our r e s u l t s i n a more accurate way. Shear and bond f a i l u r e are b r i t t l e types of f a i l u r e , there fore they have to be prevented. Thus, wa l ls [6] , [7] and [8] were constructed with a bond beam i n t h e i r upper course , and a l l of them f a i l e d i n a d i f f e r e n t mode than wal l [1] . Tdble b. I DETAILS AND RESULTS UK TESTS UN MASONRY WALL PANELS Wall Dlaenalone, Nuaber r e i n f o r c i n g , supports 1. • a 1. Maiilaua Hrt-ssure, Reinforcement rat lo P -110 pal M X -0.0011 P - « 0 psf M X Mode n l F i l l l u r r M.ijor crack pat tern at ( a l l u r e Sudden aheur and bond f a i l u r e al top courae. T H — ! i ! r T P r S r i I J Bending f a i l u r e of cant i lever port ion about v e r t i c a l r e i n -f o r c i n g . r £ r ^3 /to Centra l d e f l e c t i o n (Inches) VB Pressure (psf) Mo It* SO / .2 .4 / V .2 19 2°r f.i j r_ •" m 1 m P -160 psf •taa p u -0.0011 n Bond f a i l u r e of top bar al lowing cant i lever type f a i l u r e of top port Ion. • A f t e r removing the top course, w a l l (1) was t e s t e d a q a i n . It f a i l e d i n the same mode at a pressure of 365 psf. Table 5.1 (tont'd) P -280 par M X . p -0.0011 Rending f a i l u r e at one aide cant 11ever. "T 2oo 1*0 —r-.2 I • • • 1 • .4 * »2" 72^ 1 1 " 5 - %* • i .0011 Bending f a i l u r e of manaonry between r e i n -forcement . HoY —r— •4 fit U 12 \2V 4" ;>^!f»L«af!fs •*4 i I i i i * 4 ! P -210 max p -.0011 Aa In wall 5. P -320 psf max l> -0.0011 v P H -0.0005 Aa In walla 5 k h, v e r t i c a l reinforce-ment reached y i e l d . ESS?! Joint reinforcement CD 1 0 Table 5. I (cont'd) 4 " 4 8 ' 44" i 17 6 P -310 pal a>aa -0.0011 -0.000} Bond 111 H u n - of 14 bar nrar a ldhelght . Ver lca l re Infnrr rarn l reached y i e l d . t. 10. P -130 paf aaa P H -0.0011 P -110 paf aaa p u -0.0014 if Shear f a i l u r e along top course then bending archanlaa In b locks . Bending arcnan In In blocks . P„ -0.0002} Joint reinforcement One way bending at top changing to aerhanlna ne.ir b o H e . Table 5. I (cont'd) Y -130 pal p -0.0005 H One way bend I n * . .2 11. % P | | -0.0005 Hoatly one way bending. Very entenalve cracking. l . i . l . 1 . I T T fro •4o • P -180 paf •aa p -0.0011 v p.. -0.00025 Bending aechanlaa Very eatenalve cracking. /fit •IIIIIIIMai' I* TABLE 5.II - YIELD ANALYSIS RESULTS Wall Y i e l d Line Direction Horizontal Yield Line Equation Load at "est* Error+ "est Error** Pattern of Main Reinforcing Joint Reinforcing W = Later, load - psf My MJJ - lb . i n / i n . Large Defor-mation - psf psf psf (1) (2) (3) (4) (5) (6) (7) (8) (9) (10) 5 * " It 3fc" X v e r t i c a l none ^ H B - < V - - 7 5 M H ' = W 190 200 +5% 199 +5* 4" 04" 48 11" 7 £ " I' L" none none 210 60 200 61 -5* +2% 199 60 -5% 0% 6 10 K horizontal 5 i 2 7(M v +2M H) - W 11 48 it horizontal «vr* 5TBT< MV + 0- 5 MH ) = W 83 83 88 +6* 41 ; \ 0 45* 12 n 1.83 n 130 125 -4% 130 0% 13 7 same as 5 n v e r t i c a l tl tl n 77TO ( MV +' 7 5 MH ) = W 120 320 125 411 +4% + 28% 130 410 +8* +28% 14 44 A \ v e r t i c a l ~*—Top course sheared off ^ ( M V + . S M H ) = W 180 151 -16% 154 -14% 4 " n •M^ - v e r t i c a l moment capacity of j o i n t reinforcement spaced at 16"o.c. (p = .00025) l b . i n / i n . tW from equation in c o l . (5) using M v b = 460 or 190 l b . , M v r = 333 l b . and ^ = 0.33 t\y/h **W " " - " M v b = 510 or 235 l b . , M y r = 333 l b . and Mj, = 0.167 M v b TABLE 5.III Shear Stresses - Wall [1] - Max. Shear Wall Max. Load Stress AH V 3 - 5 / V V 156 psi (la) 310 psf (14.8 kPa) 130 psi (0.38 MPa) 0 .66 0.83 (lb) 365 psf (17.5 kPa) 135 psi (0.39 MPa) 0.69 0.87 A H = effective shear area, (as shown in Fig. 5 o = compressive strength of grout (psi) yr .22) . TABLE 5.IV Bond Stresses - Wall [1] Wall Load Stress u(psi) u u c r (la) (lb) 310 psf (14.50 kPa) 365 psf (17.5 kPa) 500 (3.45 MPa) 530 (3.66 MPa) 0.60 0.64 u = V/Z 0jd V = shear force £ 0 = sum of reinforcing bars perimeter jd = lever arm for bending internal forces u c r = ultimate bond stress = l l V f^/d b (psi) f c = taken as a g r = compressive strength of grout (psi) dj,' = bar diameter 9 4 . wood react ion wall rotary potentiometer a i r bag wooden support bear (a) V e r t i c a l l y Spanning Wall .anchors at 8 i n (200 im) t e f l o n base pads (b) H o r i z o n t a l l y Spanning Wall Figure 5.1 Test Arrangement 1 i 1 1 — o.o O.4 O.E 1.2 i.e 2.0 2:< DPL i n IN. Figure 5.2 Central Displacement V B Load - Wall (14) DPL i n IN . F igure 5.3 C e n t r a l Displacement v s . Load - Wal l (9) 96. in vc 0, ^ Q n 9. 4 o a o 6 o o o.'oo 0.'08 o7l6 0.32 0.40 0.24 DPL i n IN. Figure 5.4 Central Displacement vs. Load - Wall (10) 0.46 o o-l •D 9 o o-l o 0.00 1 0.04 ~1 0.08 0.12 DPL i n IN —r~ 0.16 0.20 Figure 5.5 Central Displacement vs. Load - wall (11) 9 7 . o s"-l o DPL i n IN. Figure 5.6 Central Displacement vs. Load - Wall (12) o o DPL In IN. Figure 5 .7 Central Displacement va. Load - Wall (13) 9 8 . o 10 o o d —I— 0.08 D i s p l a c e m e n t 0.00 —I— 0.16 1— 0.24 1 0. 32 0.4C DPL i n IN. Figure 5.8(a) Displacement of the Centre Point of the K a i l r e l a t i v e to the top and bottom displacements. s i Figure 5.8(b) Displacement of the centre point of the wal l r e l a t i v e to the edge displacements. Figure 5.B Centra l po int displacement i n a hor i zonta l and i n • . v e r t i c a l plane (wall 10). 99. e s e a c 5 e - d e e e e 1 0 . 0 0 » 1" 1 \ > » 0 . 0 4 o.oe DPL i n IN C.12 'Oisplacener.t o.ie Figure 5.9(a) Displacement of the centre point of the wal l r e l a t i v e to the top and bottor displacement. IT. « •c rz C C c — c c c . o 0.00 o!o4 O.O'B I 0.12 DPL i n IN C i s c i a c e T . e - 1 0 . 1 6 0 . 2 0 Figure 5.9(b) Displacement of the centre point of the wall r e l a t i v f to the edge displacements. Figure 5.9 C e n t r a l Point Displacement i n a Horizonta l and i n a V e r t i c a l Plane (Wall 11). psf 120-1 I n i t i a l Loading 40 •( .JO .40 .10 .10 .0) . » .»o inches Secondary Loading • • • • • • F igure 5.10 Wall (11) Load v s . Cent ra l Displacement fo r one Complete C y c l e . o T r i i i ' 0.0 0.04 0.08 0.12 0.16 0.20 0.24 STRAIN f — 1 o Figure 5.11 I.oad-Strain P lot for R e i n f o r c i n g S t e e l in Wall (4) (Test r e s u l t s for two bars p l o t t e d ) Figure S.12 Load-Strain Plot for Reinforcing Steel in Wall (6) (Test r e s u l t s for two bars plotted) o o 0.00 0.02 0.04 0.06 0.08 0.10 S T R A I N ( 1 0 _ 1 ) Figure 5.13 Load-St ra in P lo t for Re in fo rc ing Stee l i n Wall (7) (Test r e s u l t s for one bar p lo t ted) O 0) ft •a s 10 U ai • P (0 0.0 0.02 ~ i r 0.04 0. t = 7.625 i n H = B = 96 i n A s = 0.88 i n 2 f m = 2,000 p s i E m = 1,000 f m E s = 29,000,000 p s i 0.06 STRAIN (10 _ 1 ) 0.0B 1 0.1 0.12 o Figure 5.14 Load-St ra in P l o t fo r Re in forc ing S tee l i n Wall (8) (Test r e s u l t s f o r one bar p lo t ted ) 1 .2. STRAIN (10~ 2) F i q u r e "3.15 r.oad-Strain Plot for Reinforcing Steel in Wall (14) (Tost r e s u l t s for one bar plotted) o Figure 5.16 Load-St ra in P lo t f o r Re in forc ing S tee l i n Wall (9 ) (Test r e s u l t s f o r three bars p lo t ted ) 0.00 0.02 0.04 0.06 0.08 0.10 STRAIN (10~ /) Figure 5.17 Load-Strain plot for Reinforcing Steel in Wall (10) (Test r e s u l t s for two bars plotted) o 108. i rn shear area T Figure 5.18 Mechanism of F a i l u r e f or Masonry K a l i Spanning H o r i z o n t a l l y . 1 ^ 4 - 1 1 4 V % Figure 5.19 Shear Stress D i s t r i b u t i o n on Shear Area L J L v J I in Figure 5.20 Section i n V e r t i c a l Bending (a) Contact Area Reduced to a Line (b) Force Concentration Around Webs High compression (c) Effects on vertical compression due to rotation of the blocks. Figure 5.21 Horizontal Bending Mechanism Figure 5.22 E f f e c t i v e Shear Area for Out -Of -Plane F o r c e s . 111. CHAPTER 6 Summary and Conclusions In t h i s t h e s i s a study of the behaviour of masonry wal ls under com-bined in -p lane and l a t e r a l loading was presented. I t inc luded a d i s c u s s i o n of prev ious research work and the approaches of present codes of p r a c t i c e . The p roper t i es of masonry components and assemblages were examined, and the t e s t s aimed at determining the p roper t i es of the mater ia ls for the wal ls tes ted at the U n i v e r s i t y of B r i t i s h Columbia were repor ted . An e x p e r i -mental program on l a t e r a l l y loaded non- loadbearing wal l panels has been d e s c r i b e d , which w i l l eventua l ly lead to the b a s i s fo r the design of the spacing for the main s t e e l and the appropr iate amount and spacing of the d i s t r i b u t i o n s t e e l . F indings based on the r e s u l t s of the experimental program and t h e i r a n a l y s i s , can be summarized as f o l l o w s . (1) Based on the r e s u l t s of the tes ts the present code requirements on spacing seem to be very conserva t ive . (2) The wal l specimens showed a much greater strength than i s requi red for earthquake code forces app l ied to e ight foot (2.44 m) spans such as were t e s t e d . (3) J o i n t reinforcement appears to be e f f e c t i v e as d i s t r i b u t i o n s t e e l for v e r t i c a l re inforcement , and a lso as main s t e e l fo r hor i zon ta l spans. I t appears to provide moment c a p a c i t i e s corresponding to a s t ress near ly equal to i t s y i e l d s t rength . (4) L o a d - s t r a i n curves fo r the main s t e e l bars i n wal ls spanning v e r t i -c a l l y showed good agreement between current a n a l y s i s methods and experimental da ta . 112. (5) Most wal l panels f a i l i n g i n f lexure showed d u c t i l e behavior . Y i e l d l i n e a n a l y s i s , which can be j u s t i f i e d because of the observed d u c t i l e behaviour of the w a l l s , appears to be able to p r e d i c t the l a t e r a l load capaci ty of masonry to span between l i n e s of support or re inforcement . However, the moment capac i ty of the masonry i s h igh ly dependent on i n -plane loads which a r i s e from supports or c o n f i n i n g s t e e l , making an exact moment capac i ty p r e d i c t i o n very d i f f i c u l t . L a t e r a l loads pre -d i c t e d fo r moment capac i ty r a t i o s M /M of 3 and 6 d id not show any vb H s i g n i f i c a n t d i f f e r e n c e . However a f r i c t i o n a n a l y s i s seems to i n d i c a t e that a moment capac i ty r a t i o of 3 i s a more f e a s i b l e assumption. (6) The code procedures to determine the p roper t i es of masonry mater ia ls and assemblages have been shown, i n other research programs, to be unsa t i s fac to ry i n reproducing the proper t ies of masonry wal ls and t h e i r components. Although some v a r i a t i o n s i n these procedures were introduced in the mater ia l t e s t s at the U n i v e r s i t y of B . C . , fu r ther research i s necessary i n order to draw some conclus ions as to t h e i r adequacy. Further research i n t h i s program i s being aimed at determining the behaviour of masonry wal ls under ou t -o f -p lane q u a s i - s t a t i c c y c l i c l o a d i n g . The r e s u l t s obtained i n t h i s second stage of the U n i v e r s i t y of B r i t i s h Columbia research program w i l l lead to a f i n a l phase in which masonry wal ls w i l l be tested under simulated earthquake load ing on a shaking t a b l e . 113. REFERENCES 1. 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Nat ional B u i l d i n g Code of Canada, NRCC, No. 15555, Ottawa, Onta r io , Canada, 1977. 29. Canadian Standard A s s o c i a t i o n , CSA Standard A .165 .1 , Concrete Masonry U n i t s . 30. American Society for Tes t ing and Mater ia ls ASTM-C90 Standard S p e c i -f i c a t i o n s fo r Hollow Load Bearing Concrete Masonry U n i t s . 31. Roberts, J . J . , "The E f f e c t Upon the Indicated Strength of Concrete Blocks i n Compression o f Replacing Mortar with Board Capping", Proceedings F i r s t Canadian Masonry Symposium, Ca lgary , A l b e r t a , Canada, June 7-10 1976, pp . 22-39. 32. R id inger , W., Noland J . L . and Feng C . C . , "On the E f f e c t of In ter face Condi t ion and Capping Conf igura t ion on the Resul ts of Hol low-Clay Masonry Unit Compressive T e s t s " , Proceedings Second Canadian Masonry Symposium, Ottawa, Canada, June 9-11, 1980, pp. 25-39. 33. 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