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Evaluation of seismic assessment procedures to predict liquefaction and deformations Richter, Kevin J. 1995

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E V A L U A T I O N OF SEISMIC ASSESSMENT PROCEDURES T O PREDICT LIQUEFACTION A N D DEFORMATIONS  by K E V I N J. RICHTER  B . A . S c , The University of British Columbia, 1987 A THESIS SUBMITTED IN PARTIAL F U L F I L L M E N T O F T H E REQUIREMENTS FOR T H E D E G R E E O F M A S T E R OF APPLIED SCIENCE in T H E F A C U L T Y O F G R A D U A T E STUDIES Department of Civil Engineering  We accept the thesis as conforming Ho the rggSifed standard  T H E UNIVERSITY OF BRITISH C O L U M B I A April 1995 © Kevin J. Richter, 1995  In presenting this thesis in partial fulfillment of the requirements f o r an advanced degree a t the U n i v e r s i t y of B r i t i s h Columbia/ I agree t h a t the L i b r a r y s h a l l make i t f r e e l y a v a i l a b l e f o r r e f e r e n c e and study. I f u r t h e r agree t h a t p e r m i s s i o n f o r e x t e n s i v e copying of t h i s t h e s i s f o r s c h o l a r l y purposes may be granted by the head of my department or by h i s or her representatives. I t i s understood t h a t copying or p u b l i c a t i o n of t h i s t h e s i s f o r f i n a n c i a l g a i n s h a l l not be allowed without my written permission.  Department o f _ The U n i v e r s i t y of B r i t i s h Columbia Vancouver, Canada Date  11  ABSTRACT This thesis involved the seismic assessment of the foundations for a number of bridge underpass structures located in the Fraser Delta. The specific structures were selected because they had been identified by M o T H (Ministry of Transportation and Highways) as a concern due to the presence of foundation soils that could liquefy in the event of a major earthquake. Specifically, the seismic assessment involved analyzing the predicted effects of earthquake ground motion generated by 1:100 and 1:475 year events. The soil parameters required for the assessment are the pre-liquefaction shear moduli, damping parameters, as well as the soil liquefaction resistance and the post liquefaction stressstrain curves and residual strengths. These soil values were obtained using indirect test methods and experience.  With this information, several procedures were employed to assess the likely  behaviour of the structures for the design earthquakes considered. In the event of an earthquake where liquefaction is not triggered, ground movements are small.  However, when liquefaction is triggered, large deformations develop.  Thus, the first  step was to assess the liquefaction triggering potential of the foundation soils, using the soil's liquefaction resistance, shear modulus and damping values.  This liquefaction assessment was  accomplished using both total and effective stress methods. Results indicated that for the 1:100 and 1:475 year events, soil layers at the Blundell, Ladner, Mathews Road and Tsawwassen sites would liquefy.  However, no liquefaction was  indicated at the Steveston site for either the 1:100 or 1:475 events. The total and effective stress methods were in agreement in their predictions for soil liquefaction. Once it was determined that liquefaction would be triggered, the next step was to  Ill  determine the amount of deformation which would be created due to the three possible displacement mechanisms: flow sliding, lateral spreading and settlement.  Displacement  prediction required the determination and use of appropriate post liquefaction residual strengths and stress-strain curves. If a flow slide was predicted, then no further assessment was required because displacements would be very large. However, if a flow slide was not predicted, then lateral spreading and settlement potentials were assessed using a number of both analytical and emperical procedures. Test results indicated that flow slides would only be triggered at the Tsawwassen site. Eventhough flow slides were not triggered at the other sites, the potential of significant deformation was still possible as a result of lateral spreading or settlement. Remediation involving densification was examined at the Tsawwassen site.  The  examination involved varying both the amount and location of densification. The study indicated that densification directly beneath the embankments was the most successful at limiting deformations.  iv TABLE OF CONTENTS page Abstract  ii  Table of Contents  iv  List of Figures  vii  List of Tables  ix  Acknowledgement  xii  CHAPTER 1  INTRODUCTION  1  CHAPTER 2  METHOD OF ANALYSIS  3  Introduction  3  SEISMIC LOADING  24  Peak Acceleration and Velocity  24  Earthquake Time Histories  25  SOIL PROPERTIES  30  Introduction  30  Pre-Liquefaction Soil Properties  31  Liquefaction Resistance  36  Post-Liquefaction Soil Behaviour  37  CHAPTER 3  CHAPTER 4  CHAPTER 5  SITE LOCATIONS, TOPOGRAPHY, VEGETATION, AND GEOLOGY  40  Introduction  40  Site Locations  40  Site Locations - Steveston Underpass  40  Site Locations - Blundell Underpass  40  Site Locations - Ladner Exchange  43  Site Locations - Mathews Road Underpass  43  Site Locations - Tsawwassen Overpass  43  Ground Topography and Vegetation  43  Ground Water Conditions  44  Surficial Geology - General  44  Surficial Geology - Steveston Underpass Soil Conditions  45  Surficial Geology - Blundell Underpass Soil Conditions  46  Surficial Geology - Ladner Exchange Soil Conditions  46  Surficial Geology - Mathews Road Underpass Soil Conditions  49  Surficial Geology - Tsawwassen Overpass Soil Conditions  52  L I Q U E F A C T I O N POTENTIAL A N A L Y S E S  59  Liquefaction Potential Analyses - 1D-LIQ  59  Liquefaction Potential Analyses - S H A K E  61  Liquefaction Potential Analyses - S H A K E - Preliminary  62  Liquefaction Potential Analyses - S H A K E - Detailed  65  CHAPTER 7  F L O W SLIDE STABILITY A N A L Y S E S - X S T A B L  71  CHAPTER 8  DEFORMATION - DETAILED ANALYSES  79  Deformations - Horizontal  79  Deformations - Vertical  91  CHAPTER 6  CHAPTER 9 REMEDIATION MEASURES C H A P T E R 10  CONCLUSIONS A N D F U T U R E R E S E A R C H Conclusions Future Research  REFERENCES Appendix A - Soil Records - Ladner Exchange (SCPTU Records) Appendix B - Soil Records - Tsawwassen Overpass (SCPTU Records) Appendix C - Liquefaction Assessment - 1D-LIQ Appendix D - Liquefaction Assessment - S H A K E - Preliminary Appendix E - Liquefaction Assessment - S H A K E - Detailed Appendix F - Printouts from the F E M Deformation Analysis  vii LIST OF FIGURES page 1  Seismic Assessment Flow Chart  2  Relationship Between Cyclic Stress Ratio Causing Liquefaction and  4  (N^go Values for Earthquake Magnitude 7.5  7  3  Relationship Between a and K  9  4  Relationship Between Effective Vertical Stress and K,,  5  Newmark Model of A) A Block on an Inclined Plane and B) The Associated  a  9  Rigid Plastic Behaviour  17  6  Extended Newmark Model Illustrating The Work-Energy Principle  20  7  Area Plan - MoTH Site Locations  41  8  Location Plan - MoTH Site Locations  42  9  XSTABL Results - Preliminary Analysis: North Tsawwassen Site (Soil Profile Based on Old SPT Information)  10  XSTABL Results - Preliminary Analysis: South Tsawwassen Site (Soil Profile Based on Old SPT Information)  11  75  XSTABL Results - Preliminary Analysis: Blundell Site (Soil Profile Based on Old SPT Information)  13  74  XSTABL Results - Preliminary Analysis: Ladner Site (Soil Profile Based on Old SPT Information)  12  73  77  XSTABL Results - Preliminary Analysis: Mathews Road Site (Soil Profile Based on Old SPT Information)  78  Vlll  14  Geometric Configuration of the Soil Layers and Embankment Tsawwassen (Densified Soil Zone 41 - 57 m and 93 - 109 m: Test 1)  15  Geometric Configuration of the Soil Layers and Embankment Tsawwassen (Densified Soil Zone 41 - 61 m and 89 - 109 m: Test 2)  16  86  Geometric Configuration of the Soil Layers and Embankment Tsawwassen (Densified Soil Zone 51 - 65 m and 85 - 99 m: Test 4)  18  85  Geometric Configuration of the Soil Layers and Embankment Tsawwassen (Densified Soil Zone 37.5 - 61 m and 89 - 112.5 m: Test 3)  17  83  88  Geometric Configuration of the Soil Layers and Embankment Tsawwassen (Densified Soil Zone 51 - 69 m and 81 - 99 m: Test 5)  90  ix  LIST OF TABLES page 1  Table of Correction Factors for Different Magnitude To Be Used In Conjuction With The Seed Diagram Used To Determine CRR  2  6  Probability of Various Earthquake Generated Ground Motion (North Side of the George Massey Tunnel)  25  3  Probability of Various Earthquake Generated Ground Motion (Ladner)  25  4  Preliminary Assessment Earthquake Records Used  27  5  Detailed Assessment Earthquake Records Used  29  6  Table of Proposed Shear Modulus Values for Various Soil Types  34  7  Summary of Available Recommended Maximum Damping Ratios for Various Soils  8  35  Strength Coefficient Table Based on (Nl)60 Values and Limiting Strains  38  9  Correction ( A N ^ Values Due to the Presence of Silt  38  10  Summary of Estimated Soil Parameters - Steveston Underpass  47  11  Summary of Estimated Soil Parameters - Blundell Underpass  48  12  Summary of Estimated Soil Parameters - Ladner Exchange  50  13  Summary of Estimated Soil Parameters Used in the Detailed SHAKE Analyses - Ladner Exchange  51  14  Summary of Estimated Soil Parameters - Mathews Road Underpass  53  15  Summary of Estimated Soil Parameters - Tsawwassen Overpass North  54  X  16  Summary of Estimated Soil Parameters - Tsawwassen Overpass South  17  Summary of Estimated Soil Parameters Used in the Detailed SHAKE Analyses - Tsawwassen Site  18  61  Summary of Liquefiable Zones for the 1:100 Year Event - Preliminary Soil Profile Tsawwassen North (SHAKE)  26  61  Summary of Liquefiable Zones for the 1:475 Year Event - Preliminary Soil Profile Ladner Exchange (1D-LIQ)  25  60  Summary of Liquefiable Zones for the 1:100 Year Event - Preliminary Soil Profile Ladner Exchange (1D-LIQ)  24  60  Summary of Liquefiable Zones for the 1:475 Year Event - Preliminary Soil Profile Tsawwassen South (1D-LIQ)  23  60  Summary of Liquefiable Zones for the 1:100 Year Event - Preliminary Soil Profile Tsawwassen South (1D-LIQ)  22  60  Summary of Liquefiable Zones for the 1:475 Year Event - Preliminary Soil Profile, Tsawwassen North (1D-LIQ)  21  58  Summary of Liquefiable Zones for the 1:100 Year Event - Preliminary Soil Profile, Tsawwassen North (1D-LIQ)  20  57  Summary of Estimated Soil Parameters Used in the Finite Element Analyses - Tsawwassen Site  19  55  62  Summary of Liquefiable Zones for the 1:475 Year Event - Preliminary Soil Profile Tsawwassen North (SHAKE)  62  XI  27  Summary of Liquefiable Zones for the 1:100 Year Event - Preliminary Soil Profile Tsawwassen South (SHAKE)  28  Summary of Liquefiable Zones for the 1:475 Year Event - Preliminary Soil Profile Tsawwassen South (SHAKE)  29  64  Summary of Liquefiable Zones for the 1:100 Year Event - Preliminary Soil Profile Mathews Road Underpass (SHAKE)  32  63  Summary of Liquefiable Zones for the 1:475 Year Event - Preliminary Soil Profile Ladner Exchange (SHAKE)  31  63  Summary of Liquefiable Zones for the 1:100 Year Event - Preliminary Soil Profile Ladner Exchange (SHAKE)  30  63  64  Summary of Liquefiable Zones for the 1:475 Year Event - Preliminary Soil Profile Mathews Road Underpass (SHAKE)  64  33  Summary of Residual Strengths Used in Flow Slide Assessment  72  34  Summary of Predicted Horizontal Deformations Based on S implied and  35  Detailed Analysis Methods  82  Estimated Average Settlements Based on the Tokimatsu and Seed Procedure  93  xii  ACKNOWLEDGEMENTS  The time taken to complete this thesis has been long. Had it not been for the persistent support of the special few, these following words of gratitude may not have been necessary. To Dr. P. Byrne, I am grateful for his guidance in choosing a thesis topic and his influence, support and perseverance in working through of this thesis. Dr. T. Ersoy, who provided a multiplicity of support which was never asked for but was always graciously dispensed.  For your constant positive support, limitless patience and  enthusiasm for this thesis, I thank you for everything and owe you much more than just these few pages. I would also like to thank Mr. R. Lidgren, for just being who he is in helping this thesis come to fulfilment. To the Ministry of Transportation and Highways of British Columbia for providing the much appreciated financial assistance which made all this work possible. Finally, Laura E . Richter, who had to put up with so much and get so little in return. I acknowledge you for your proof-reading, editing, discussing, listening, waiting and supporting in more ways than I could have imagined. Therefore, from my heart and soul, I thank you, love you and dedicate this thesis to you.  1  CHAPTER 1 INTRODUCTION The problem investigated in this thesis is the seismic response of five bridge overpass structures in the Fraser Delta which are underlain by loose saturated soils that could liquefy during 1:100 or 1:475 year earthquake events. During a major earthquake, damage to structures can be the result of ground shaking and/or deformation due to liquefaction. Structures must be capable of resisting or remediation options must be instituted to mitigate the damage. The seismic response of a soil-structure depends upon the seismic input at the base rock level, the geometry and properties of the overlying soils, and the soil-structure. In dealing with loose saturated soils, a major concern is liquefaction wherein the soil loses its stiffness and strength under cyclic loading. The objective of this thesis is to predict the extent of liquefaction and the response in terms of strains and displacements of the soil structure. This involved: 1)  The earthquake input motions on firm ground beneath the base of the  structure. 2)  The geometry, pre-liquefaction stress-strain properties, properties of the  various soil zones within the soil structure. 3) Dynamic analyses of the soil structure to compute the cyclic stresses induced by the design earthquake. 4) Assessment of liquefaction by comparing the dynamic stresses caused by the design earthquake with the cyclic resistance of the soil. 5) Assessment of flow slide potential.  2  6) Assessment of post liquefaction deformations. 7) Evaluation of different remediation options which attempt to mitigate the post liquefaction deformations. The following is a brief outline of the chapter organization for this thesis. Chapter two involves a comprehensive discussion of the method of analysis used in the seismic assessment. Chapter three provides a discussion of the design seismicity and earthquake time histories utilized. Chapter four provides a discussion of how the necessary critical soil parameters and properties were obtained. Chapter five presents the information pertaining to the site locations, topography, vegetation, and geology/soil conditions. In chapter six, the liquefaction potential at all five sites is assessed. This chapter includes a comparison between the total and effective stress methods of predicting the onset of liquefaction. Chapter seven provides an assessment of the development of flow slides. In chapter eight, the horizontal and vertical deformations associated with lateral spreading and settlement are presented for the Tsawwassen site. In addition, this chapter summarizes the effects of remediation on the reduction of deformations. In chapter nine, the advantages and disadvantages  of several different types of  remediation possibilities are discussed. Chapter ten presents the conclusions and future research.  3 CHAPTER 2 METHOD OF ANALYSIS INTRODUCTION During an earthquake, the soil is subjected to cyclic loading which causes soil particles to move within the skeleton, resulting in an accumulation of volumetric strains if drained. If the pore spaces within the skeleton are filled with water and the water is prevented from escaping, then upon shaking a transfer of normal load from the skeleton to the water occurs resulting in a rise in pore water pressure and a drop in stiffness and strength.  At the point  where the pore pressure equals the overlying normal stress, liquefaction is triggered.  Upon  liquefaction, the strength and stiffness of the soil is dramatically altered, resulting in significant deformations.  These deformations can take the form of flow slides, lateral spreading or  settlement and can be very significant. In order to predict the response of soil during an earthquake, it is necessary to conduct a seismic assessment. For this thesis an uncoupled dynamic assessment was performed which included total and effective stress methodologies. This chapter describes in detail the procedures that were employed to conduct the analyses.  In order to assist in understanding how a seismic  assessment is conducted, a flow chart (refer to figure 1) has been prepared which illustrates the order of steps that are taken. The first analysis procedure involved evaluating the liquefaction triggering potential of the soil layers at each site. This was accomplished by utilizing both total and effective stress analysis procedures. The total stress approach employed a total stress equivalent elastic dynamic analysis to determine small strains prior to the triggering of liquefaction (Byrne, P., et. al.,  SOIL PARAMETERS STRESS-STRAIN RELATIONS GEOMETRY  ( C R R , Sr, P o s t - L i q R e s p o n s e )  ANALYSES  LIQUEFACTION ASSESSMENT (CRR)  YES  NO  LIQUEFACTION  LIQUEFACTION  FLOW SLIDE ASSESSMENT (Sr)  NO  YES FLOW SLIDE  FLOW SLIDE  LATERAL SPREADING  SETTLEMENT  ASSESSMENT  ASSESSMENT  (Post-Liq Response)  (Post-Liq Response)  Figure 1 Seismic Assessment Flow Chart  5 1993). The effective stress approach modelled the complete stress/strain/pore pressure response during the seismic loading. In the total stress approach, the cyclic shear stress ratio (CSR) generated by the input earthquake time histories was compared with the corrected cyclic resistance ratio (CRR ) of the 1C  soil layer which is based on strain development criteria. A factor of safety against the triggering of liquefaction was determined (refer to equation 1). Generally, a factor of safety of 1.1 to 1.3 is considered reasonable (Byrne, P., 1991). FSL = CRR /CSR  eqn. 1  1C  The cyclic shear stress ratio is defined as, CSR = 0.65 (r ) dy  max  / a\ where  (T ) d y  m a x  is the  maximum shear stress calculated at the middle of the desired soil layer, <r' is the effective 0  overburden stress at the middle of the desired soil layer assuming the effects of a rising pore pressure are not included, and the factor 0.65 is used to convert from a single random maximum to an equivalent uniform shear stress (Seed, H . , et. al., 1983). The cyclic shear stress ratio for each soil layer is computed using the one dimensional computer program SHAKE. As a result of SHAKE being a one dimensional program, it lacks the ability to evaluate two and three dimensional effects on the development of liquefaction. When cyclic shear stress results were computed at various locations in an embankment using the one dimensional approach (SHAKE) with a two dimensional approach (2-D FLUSH), it was noted (Jong, H., 1988) that when the cyclic shear stresses were compared, the peak cyclic shear stresses calculated using the one dimensional approach were 5 to 15 % lower than the stresses generated using the two dimensional approach. The cyclic resistance ratio depends upon the in-situ conditions of the soil. Specifically,  6 such factors as density, gradation, fines content, plasticity, confining stress and static bias influence the resistance of the soil against liquefaction. The best way to determine the cyclic resistance ratio is direct testing of undisturbed samples.  However, due to the difficulty and  associated cost, typically indirect methods based on SPT ( N ^ values (standard penetration tests corrected for overburden and hammer energy) and field experience are used. For this thesis, determination of the cyclic resistance ratios were based on actual ( N ^ or SCPTU (seismic cone penetrometer test) correlated to SPT ( N ^ values.  This was accomplished by using Seed's  Liquefaction Assessment Diagram (refer to figure 2) which correlates empirically the CRRj with the (NJgo values based on existing historical liquefaction experience (Pillai, V.S., and Stewart, R.A., 1993). Seed's diagram is for an earthquake magnitude of 7.5, 15 cycles of strong ground motion and percentage of fines in the soil. Therefore, in order to use this diagram for different magnitude earthquakes and associated different number of cycles, appropriate correction factors must be applied and are shown in table 1 (Seed, et. al., 1983). T A B L E 1.  Table of Correction Factors for Different Magnitude To Be Used In Conjunction With The Seed Diagram Used to Determine CRR (Seed, et. al., 1983)  Earthquake Magnitude M  # of Cycles at 0.65 r  max  {(CRR)  M=M  /(CRR)  8.5  26  0.89  7.5  15  1.0  6.75  10  1.13  6  5  1.32  5.25  2-3  1.5  M=75  }  To determine the correction factor (K ) for earthquake magnitudes other than those given M  Figure 2 Relationship Between Cyclic Stress Ratio Causing Liquefaction and (N^Vaiues for Earthquake Magnitude 7.5 (Seed, H., et. al., 1984)  8  by this table, interpolation between the provided values can be made.  For example, for a  magnitude 7 earthquake, a correction factor of 1.09 could be used. During the stages of text revision for this thesis, further outside research (Loertscher, T., and Youd, T., 1994) indicated that the K values used in the preparation of this thesis may have M  been too conservative.  If the new K values are used, it is anticipated that the probability of M  liquefaction at the sites would be reduced. Consequently, re-evaluating these sites using the new K values is recommended. M  In addition, the effects of static bias and confining stress must be taken into account in the determination of the corrected CRR value. The effects of static bias (static driving shear stresses) are a concern because in loose soils, a static bias can decrease the resistance of the soil to the initiation of liquefaction (Seed, R., and Harder, L . , 1990). A correction factor (KJ can be applied to the CRR^ in order to account for static bias. Based on available information, the following figure has been created to provide a value for K (refer to figure 3). Typically, two a  dimensional finite element programs are employed to calculate the stress state for the soil if level ground conditions do not exist. Confining stress is also of concern in predicting soil liquefaction as field evidence has shown that confining stress level influences the triggering of liquefaction (Seed, R., and Harder, L . , 1990). As the confining stress increases, the normalized cyclic resistance is reduced. The effects can be accounted for by obtaining a correction factor from figure 4. Once CRR  l5  K , K„ and K have been determined, a final corrected CRR M  obtained (refer to equation 2).  a  CRR  1C  1C  can be  is then used in the calculation of the factor of safety  against liquefaction (refer to equation 1).  :  1  0  1  O.I  0.2  J  I  0.3  0.4  0.5  Figure 3 Relationship Between°<and (Seed, R., and Harder, L, i990) ~4  J  •o  1.0  \  •  •  o  o  0.8  ^  -w  a  <;  •  0.6  ^ •  FAIRMONT O A U  •  LAME  ARR0WH£ AO  SHEFFIELO  OAM  0  UPPER  LEANOPO  A  UPPER  A  LQS  •  PERB'S  •  SAROlS O A M S M E L L  SAN  O  0 . 2  S"£LL  Q  SAN FESNANOO  SAROLS  OAM  OAU  SHELL  OAM  SHELL  •  SHELL  OAM S H E L L .  «C'93,100%  O A M FOUNDATION  •  TMERUALITO  •  THERMALIFO  Q  ANfELOPE  <•>  FORI  PEC*  AFTERBAT  OAU  fO**CBA*  O A U IMPERVIOUS OAM  SACRAMENTO  RIVER  SANO,  MONTEREY  SANO,  0»  RElO B E G F O P O  0  NEw  JCSSEY  1 . 0  FOUNDATION  UAIERIAL  SHELL  •  0  T  F0UN0*I<0N  OAU  a  0  SHELL  OAM  SAN FERNANDO  ANGELES  a  i  OAU  V  LO<*ER  0 . 4  i  OI  = 38.  60,  ra,  100%  =40%  S A N O , Or - 40, 6 0 % BACKFILL,  2  FPl  .  E F F E C T I V E  flC»9i%  0  3  .  0  C O N F I N I N G  4  .  0 5  P R E S S U R E  .  0  6 . 0  7 . 0  8.0  ( t s f ) o r ( k s c )  Figure 4 Relationship Between Effective Vertical Stress and K^Seed, R., and Harder, L, 1990)  10 CRR  1C  = CRRj * K * K * K a  M  ff  eqn. 2  In the total stress approach, the dynamic response of the soil is simulated by a damped non-linear elastic model which considers a reduction in soil stiffness due to shear strains only (Finn, W., et. al., 1976). The total stress analysis is thus limited because it does not recognize that as the soil is strained, pore pressures rise causing a reduction in effective stress. The total stress analysis does not take into account this decrease in effective stress on the soil stiffness. As an alternative, an effective stress approach can be employed to predict the liquefaction triggering potential of soil. In this approach, the effects of seismically induced pore pressures on stiffness and strength can be taken into account continuously during the analysis (Finn, W., 1988). During earthquake loading, the soil structure is subject to several cycles of loading and unloading which results in non-recoverable plastic strains. These plastic volumetric strains cause a rise in pore pressure when drainage is restricted and contributes to changes in the non-linear stress-strain behaviour of the soil.  Consequently, the effective stress approach models the  response of the soil more accurately than the total stress approach when it is subject to several cycles of dynamic seismic loading. For further information on different effective stress models refer to Byrne, P., and Yan, L . , 1990, Byrne, P., 1990b, Finn, W., 1988, Finn, W., et. al., 1977, and Finn, W., et. al., 1976. There are a number of computer codes which perform an effective stress dynamic analysis. One such program is 1D-LIQ. In this program, a one dimensional soil column is used and is divided into several horizontal layers. This layered system is defined as a lumped mass system, lumping one-half of the mass of each layer at the layer boundaries. These layers are  11 represented as a series of masses and are connected to each other by dashpots and nonlinear springs (Byrne, P., and Yan, L . , 1990, Finn, W., et. al., 1977, Finn, W., et. al., 1976). The dashpots model the soil's damping characteristics such that the nonlinear behaviour is simulated. The springs model the soil's stiffness and have stress-strain relations which reflect the nonlinear and strain-dependent behaviour of the soil. Equation 3 is used to model the initial loading conditions, while equation 4 is used for subsequent unloading and reloading). T  where  =  T G y r  { G  M  O  7} / {1  +  ( G  M  O  7  / O}  shear stress at strain amplitude y initial maximum tangent modulus strain amplitude maximum shear stress that can be applied to the sand i n its initial state without failure  m o  m o  (r-r )/2 = { G ( - 7 ) / 2 } / { 1 + ( ( G J ( Y - Yj / 2 r J r  where  mn  r  eqn. 3  r  y TJUJJ  7  r  eqn. 4  loading reversal generated shear stress at strain amplitude y new maximum tangent modulus loading reversal generated strain amplitude maximum shear stress that can be applied to the sand without failure  These stress-strain equations are then incorporated into the differential equations of motion (refer to equation 5). [M]{x} + [C]{x} + [K]{x} = -[M]{u }(t) g  where  [M] [C] [K] {u } x,x,x g  eqn. 5  diagonal mass matrix damping matrix stiffness matrix base rock acceleration relative accelerations, velocities, and displacements o f the masses  To calculate the behaviour of the soil column, 1D-LIQ uncouples these equations of motion and then numerically integrates each to give the response of each mode. In the analysis,  12 it is assumed that the pore pressure is known at the start of the time period A T . Strains are then assumed and are used to compute G (modulus values) and C (damping values) for each layer. Displacements (x) can be determined at various times within A T and consequently the strains in each layer. Strains are then used to calculate displacements. These calculated strains are then used to calculate shear modulus. If the difference between the new modulus value and the old modulus value, divided by the new modulus value is less the 5%, the new modulus value is assumed to be acceptable. However, if the value is greater than 5 %, then process is repeated until compatiblity is achieved. When compatible modulus have been determined, plastic volume changes caused by the strains are computed. The increments of plastic volumetric strains are computed using equation 6 (Byrne, P., 1990b), and are then used to determine the rise in pore pressure (refer to equation 7 and 8). Ae where  = 0.5 7 C Exp(-C (e /7))  p v  t  y e  p v  C,,C  2  2  p v  amplitude o f the shear strain i n the half-cycle considered accumulated plastic volumetric strain constants that depend on the type of sands and their relative density (Byrne, P . , 1990b)  Au = M Ae  p v  u = E Au where  u M  eqn. 6  eqn. 7  eqn. 8  pore pressure constrained rebound effective stress tangent modulus  Once the pore pressure rise and associated strains are known, the shear modulus can be calculated.  For dry or drained conditions, settlements are computed from the resulting  13 volumetric strains. However, for the undrained conditions, pore pressures are computed and are used to modify the modulus for the next time interval. Included in 1D-LIQ is the option to permit the dissipation of pore pressures.  Thus, using an effective stress uncoupled iterative  approach, the shear modulus and damping values are varied with time as the strain level and pore pressures rise. After it has been determined that a soil layer liquefies either by a total or effective stress analysis, the first step is to assess whether a flow slide will develop. A flow slide assessment consists of assigning a reduced strength to the liquefied layer and then performing a limit equilibrium analysis.  Calculating the strength of the liquefied layer can be based on direct  methods such as laboratory testing (Finn, W., 1993, Salgado, F., and Pillai, V . , 1993, Poulos, S., et. al., 1985) and indirect methods such as ( N ^ values which are correlated with historical evidence (Seed, R., and Harder, L., 1990, and Seed, H . , 1987). This is discussed in detail in chapter 4. After the liquefiable zones have been identified and assigned a residual strength, a limit equilibrium analysis can be conducted. For these stability analyses, no seismic coefficient is employed. The purpose is to ascertain that after the shaking has stopped does the soil have sufficient strength to prevent a static failure. The XSTABL computer program (Sharma, S., 1990), using the Bishop method of irregular surfaces, was the limit equilibrium program employed for these analyses. Even when liquefaction is triggered, flow slides may not be initiated, but large deformations may still develop as a result of lateral spreading. This occurs because large strains are required to mobilize the residual strength and these large strains can lead to very large  14 displacements that may not be tolerable. Several methods exist which can be employed for predicting seismicly induced deformations due to lateral spreading. These methods range in their degree of complexity and simplifying assumptions employed, and can be divided into two general categories: EMPIRICAL which is based on field observations and ANALYTICAL which is based on numerical calculations. Two examples of the empirical approach include the Hamada method (Hamada, M . , et. al., 1987) and Bartlett-Youd method (Bartlett, S., and Youd, T., 1992). These two methods were based on observing past liquefication induced displacements and then modelling this behaviour with simplified equations. In the Hamada method (Hamada, M . , et. al., 1987), permanent ground displacements, which were generated by liquefaction, were measured from airphoto graphs and ground traverses of the cities Noshiro and Niigata, both located in Japan. With this information, most of which was from the Noshiro data base, a regression analysis was performed in order to derive a simplified equation which best modelled the observed displacements.  It was observed that  displacements were highly dependent upon the thickness of the liquefied layer and the slope of the ground surface. Utilizing this information, the following equation (refer to equation 9) was determined. D = 0.75 (T) where  D T S  1/2  (S)  1/3  eqn. 9  lateral displacement i n meters thickness of liquefied layer i n meters ground slope i n %  In the Barlett-Youd (Bartlett, S., and Youd, T., 1992) method, a multiple linear regression (MLR) analysis of liquefaction induced horizontal displacements was conducted. The  15 data base used in the analysis included the following earthquakes: 1906 San Francisco California, 1964 Alaska - Alaska, 1964 Niigata - Japan, 1971 San Fernando - California, 1979 Imperial Valley - California, 1983 Nihonkai-Chubu - Japan, 1983 Borah Peak - Idaho, and 1986 Superstition Hills - California. In determining the equation, numerous variables were evaluated in the regression and included such items as earthquake magnitude, topography, soil factors, and geological conditions.  The analysis indicated that two equations, free face and ground slope  condition, were required to predict horizontal displacements (refer to equations 10 and 11). Free Face Condition (movement of ground into an open space such as a channel) LOG(D +0.01) = - 16.366 + 1.178*M - 0.927*LOG(R) - 0.013*R + 0.657*LOG(W) + 0.348*LOG(T ) + 4.527*LOG(100-F ) - 0.922*D50 eqn. 10 H  15  15  15  Ground Slope Condition (movement of ground which has a slope) LOG(D +0.01) = - 15.787 + 1.178*M - 0.927*LOG(R) - 0.013*R + 0.429*LOG(S) + 0.348*LOG(T ) + 4.527*LOG(100-F ) - 0.922*D50 eqn. 11 H  15  where  D M R W S T  H  I 5  Fl5  D50  15  15  horizontal displacement (m) earthquake Richter magnitude horizontal distance from energy source (km) 100*(height {H} of free face/distance {L} from the free face) ground slope i n % cumulative thickness of saturated granular layers with ( N J ^ less than 15, (m) average fines content o f the T layer (%) average mean grain size o f the T layer (mm) 1 5  15  1 5  Both methods provide a very simple means of predicting liquefaction induced displacements. They are only crude methods which are limited in accuracy due to their lack of ability to model the soil's stiffness, strength and stress-strain behaviour as it deforms.  As a  result, these methods are limited to predicting the magnitude of the horizontal displacements. The analytical method is the other strategy utilized to predict liquefaction induced displacements and is composed of the following typical numerical methods:  Newmark  (Newmark, N . , 1965), Extended Newmark (Byrne, P., 1990c), and Complete Dynamic Effective  16 Stress - TARA-3 (Finn., W . , et. al., 1986). These methods vary in complexity, time required and thus in their associated cost, with the Newmark representing the simplest and the complete dynamic effective stress being the most complicated. Between these two extremes in complexity is the Extended Newmark method. The Newmark method simulates the behaviour of a block sliding on a inclined plane which effectively represents a single degree of freedom rigid plastic system (refer to figure 5). In this method, the residual strength of the liquefied layer (post liquefaction soil strength) provides the sliding resistance, while the driving energy is comprised of both the static and earthquake driving components. The static energy is simply the weight vector component which acts down the inclined surface. The earthquake driving energy represents an energy pulse which is applied to the block. This energy pulse is a direct result of the soil having a velocity and acceleration during the earthquake shaking. Consequently, this work energy principle assumes that the work done is equal to the change in kinetic energy which is equal to the soil resistance energy minus the driving stress energy (refer to equation 12). This equation could then be reduced to equations 13 and 14. Work Done where  1/2 * M * V 1/2MV SJ) TD M V S D T  2  r  2  = (S * D) - ( T * D) = (S -r) * D r  r  eqn.  12  eqn.  13  eqn.  14  change in kinetic energy work done by soil resistance work done by driving stress mass of sliding block yield velocity of sliding block residual strength of soil displacement of block driving shear stress  1/2 * M * V  2  = (S -r) * D r  D = {M * V } / {2 * (S -r)} 2  r  A) V = Velocity M = Mass of B l o c k D = Seismic Displacement  B)  Force Soil Resistance  Displacement, D  Figure 5 Newmark Model of A) A Block On An Inclined Plane and B) The Associated Rigid Plastic Behaviour (Byrne, P., 1994b)  18 Also knowing that the yield acceleration could be defined (refer to equation 15). N = where  N S T M g r  (S -r) / M * g  eqn. 15  r  maximum yield acceleration coefficient residual strength of soil driving shear stress mass acceleration o f gravity  Thus combining these equations, the horizontal displacement could be predicted for one pulse of energy (refer to equation 16). D = . V / {2 * g * N}  eqn. 16  2  where  N V g  maximum yield acceleration coefficient yield velocity of sliding acceleration of gravity  However, Newmark recognized that during an earthquake the block would be subject to several pulses. In order to determine the number of earthquake energy pulses, Newmark applied several actual time histories and computed the associated displacements.  From these analyses,  Newmark observed that typically six pulses of energy provided an upper bound to the predicted deformations. It must be recognized that this generalization is not always true. Consequently, a simple equation (refer to equation 17) was computed based on six energy pulses, peak velocity (V), and maximum yield acceleration coefficient of the slope (N).  The maximum yield  acceleration coefficient is determined from a limit equilibrium analysis where the maximum yield acceleration coefficient represents the seismic coefficient for a factor of safety of one. D = (6 * V ) / (2 * g * N)  eqn. 17  2  where  D V g N  horizontal ground displacement (m) peak velocity (m/s) acceleration of gravity (m/s ) maximum yield acceleration coefficient 2  19 While the Newmark method accounts for the reduction in soil strength once liquefaction has been triggered, this method modelled the soil's stress strain behaviour with a rigid plastic spring which does not accurately model the soil stress strain response. As a result, an extended Newmark method was proposed (Byrne, P., 1990c) which modelled the stiffness and residual strength of the liquefied layer with a nonlinear spring (refer to figure 6).  In this extended  Newmark method, it is recognized that at the moment liquefaction is triggered, a dramatic reduction in stiffness occurs. As a result, the liquefied soil is softened which allows the static driving stresses to move the soil. When the soil moves, the soil will have an associated velocity and consequently, the soil will move an additional amount due to these dynamic forces. The entire movement, however, will be subject to the balancing of the work done. Comparing the Newmark's stress strain curve with the extended Newmark stress strain curve, it should be noted that Newmark method neglects the displacements from P to S (Jinto, H., and Byrne, P., 1994). Strains generated between P and S could range from 20 to 50 %, depending upon the soil density, and consequently, displacements could be considerable before the soil's residual strength is achieved.  A simplified program, LIQDISP (Byrne, P., and Yan, L . , 1990), has been  developed which incorporates this method and review of model predictions with observed field and laboratory shaking table values indicate good agreement. The next level in predicting liquefaction displacements involves the use of two dimensional finite element programs which incorporate the non-linear stress strain response of the soil, the residual strength of the soil, and the dynamic forces acting on the soil. This method is also more valuable than the previously discussed displacement prediction methods because this method predicts the displacement pattern of the entire body. This method also recognizes that  Q  Strain  Figure 6 Extended Newmark Model Illustrating The Work-Energy Principle (Byrne, P., 1994b)  21 deformations are the result of an accumulation of strain increments throughout the entire body of the soil structure. Thus a realistic deformation pattern is predicted rather than a single displacement of one point which allows the prediction of displacements for complex geometries. One such computer program which employs these approaches and was used in the determination of ground deformations for this thesis was SOILSTRESS version DEF13 (Byrne, P., and Janzen, W., 1981). This program modelled the soil behaviour utilizing a nonlinear, isotropic, elastic, hyperbolic stress strain relation. The nonlinear soil behaviour was modelled with an equivalent elastic or secant moduli which is stress level dependant. This program also incorporates the extended Newmark work energy theorem (Jinto, H . , and Byrne, P., 1994). For the multi-degree-of-freedom finite element system, the displacements were calculated using a energy balance matrix equation (refer to equation 18). {K} {A} = {F + AF} where  [K] {A} {F} {AF}  eqn. 18  global stiffness matrix of the system vector of nodal displacements static load vector acting on the system (gravity plus the boundary loads) additional load vector applied to produce displacements to satisfy the energy balance for equation 19  eqn. 19 where  W .ext WM V  external work internal work mass initial mass velocity and assuming final velocity is zero  For a single degree of freedom system when {AF} = 0, the displacement of the block would be equal to R (refer to figure 6, from Jinto, H., and Byrne, P.M., 1994). Consequently, an additional force would be required to balance the energy and predict the displacement from  22 points S or T (refer to figure 6). This additional force can be created using a seismic coefficient k (refer to equation 20). {AF} = {k W} where  AF k  W  eqn. 20  additional load vector pseudo seismic coefficient which is used to balance the energy i n accordance with equation 20, and note that this is not the peak ground acceleration but is instead a random chosen number weight of the soil element  As discussed previously, liquefaction generated horizontal displacements are due to gradual strain hardening when the soil has been significantly softened due to the triggering of liquefaction and due to the kinetic energy contained in the soil created by the soil mass having a velocity at the moment of liquefaction. Looking at figure 6, it can be noticed that the portion of displacement attributed to the soft, strain hardening stage is from Q to S and the portion of displacement due to the trapped kinetic energy is from points S to T. It has been determined (Jinto, H . , and Byrne, P., 1994) that the displacement due to softening can be best computed using {AF} which is based on a vertical seismic coefficient (gravitational forces) and the displacement due to the kinetic energy can best be determined using {AF} which is based on a horizontal seismic coefficient (horizontal momentum forces). For the multi-degree-of-freedom finite element system, the W represents the work done int  by the element stresses and strains while W  ext  depicts the work done by the static load vector,  {F} * {A} (Byrne, P., et. al., 1992). Thus, the additional force {AF} is adjusted to predict T  displacements {A} and to balance the energy theorem requirements. Also, this additional force is not included when computing the work done by the static force. Results generated using this approach were compared with real deformations of the  23 observed at the upper and lower San Fernando dams and were noted to be in very close agreement (Byrne, P., et. al., 1992). Consequently, it was decided that this program provides an accurate, cost effective and rapid means of predicting liquefaction generated deformations. The next level in complexity involves the use of a complete effective stress dynamic model. This approach is a complete package analysis because it calculates the generation of excess pore water pressures within each element as the earthquake shaking proceeds. Thus, zones of liquefaction are identified and the resulting deformations predicted. But in addition, the redistribution of pore pressures are determined and can be employed to assess the potential development of delayed failures. Two such computer programs which embody this approach are TARA-3 (Finn, W., 1988, Finn, W., et. al., 1986) and FLAC-UBC (Byrne, P., 1994b). Both programs, however, are essentially research tools and are not considered appropriate analysis tools for minor structures (Byrne, P., et. al., 1992).  24  CHAPTER 3 SEISMIC LOADING INTRODUCTION Selection of appropriate earthquake time histories scaled to the desired peak accelerations and velocities is paramount in conducting a seismic assessment. Earthquake time histories are dependent on several variables and the choice of suitable earthquake records must be based on a logical selection criteria.  The selection of appropriate time histories is based on such factors  as earthquake type (e.x. strike-slip, or thrust), magnitude, signal character, duration, frequency content, distance from the site to the epicentre, focal depth and whether the signal was recorded on soil or on rock. Time histories used in an assessment can either be chosen from an expansive available earthquake record library or created synthetically. PEAK ACCELERATION AND VELOCITY In order to determine the peak acceleration and velocity for the design 1:100 and 1:475 year events, the Pacific Geoscience Centre was asked to conduct a seismic assessment in the vicinity of the George Massey Tunnel and in the town of Ladner. These two sites were chosen because they were central to all of the sites, and thus the results could be used for all of the sites. The values obtained by the Pacific Geoscience Centre for these two locations are listed in tables 2 and 3. The Task Force Report (Anderson, D . , et. al., 1991) advocated that a Richter Magnitude earthquake of M7 be utilized in the preparation of a seismic assessment. As a result, it was decided that Richter Magnitude earthquake of M7 be used in the analysis.  25 T A B L E 2.  Probability of Various Earthquake Generated Ground Motion (North Side of the George Massev Tunnel) (from the Pacific Geoscience Centre, 1989, and included in a report by Ker, Priestman regarding the George Massey Tunnel No. 1509, Response to Earthquakes, File No. 2652, November 1989, for the Bridge Engineering Branch, MOTH)  Probability of Exceedance per Annum  0.01  0.005  0.0021  0.001  Probability of Exceedance in 50 Years  40%  22%  10%  5%  Return Period (yrs)  100  200  475  1000  Peak Horizontal Ground Acceleration (G)  0.097  0.147  0.238  0.344  Peak Horizontal Ground Velocity (m/sec)  0.078  0.124  0.219  0.334  T A B L E 3.  Probability of Various Earthquake Generated Ground Motion (Ladner) (from the Pacific Geoscience Centre, 1991, and included in class notes provided in Civil Engineering Course Civil 581, 1990, University of British Columbia)  Probability of Exceedance per Annum  0.01  0.005  0.0021  0.001  Probability of Exceedance in 50 Years  40%  22%  10%  5%  Return Period (yrs)  100  200  475  1000  Peak Horizontal Ground Acceleration (G)  0.099  0.151  0.244  0.355  Peak Horizontal Ground Velocity (m/sec)  0.079  0.126  0.222  0.342  EARTHQUAKE TIME HISTORIES As discussed earlier, the selection of earthquake time histories is based on many factors. For the purpose of this thesis it was decided to choose earthquake time histories which would best represent the anticipated earthquake. The selection criteria included: Richter earthquake magnitudes between 6 and 8, A/V ratios approximately equal to one, and records which were  26 recorded on bedrock. Earthquakes with Richter magnitudes between 6 and 8 were selected because it was determined that these would be the likely magnitude of earthquakes which would be responsible for the design 1:100 and 1:475 year events. By utilizing these records, it was assumed that scaling of only the accelerations and velocities was necessary. The signal duration, frequency content and phase characteristics were considered appropriate and did not need modification. The A/V ratio (acceleration/velocity) was also considered an important factor because it represented a relative method of evaluating the distance to the epicentre from the recording station. For example, an A/V ratio less than 1 indicated that the epicentre was further away than for an A/V ratio greater than 1. Information available from the GSC indicated that the anticipated acceleration and velocity would be 0. lg and 0.1 m/s respectively. As a result, this information yielded an A/V ratio of 1 for these selected sites. Next, it was considered important to use only records which had been recorded on bedrock. This was considered important because it is known that earthquake signals are altered as they propagate through soil. The frequency content, signal amplitude and signal duration are some of the record characteristics that are effected (Telford, W., et. al., 1976). Therefore, to eliminate the inappropriate records, only records obtained on bedrock were used. In the development of this thesis, the work was divided into two phases, a preliminary assessment phase and a detailed evaluation phase. For the preliminary phase, four earthquake records from the San Fernando earthquake of 1971 were used in analyzing all five sites. This earthquake was a magnitude 6.4 on the Richter scale. The records used were the Griffith Park, Cal Tech, Lake Hughes Array Station #4 and Lake Hughes Array Station #12 (refer to table 4).  27 These records were chosen because, at the time of the preliminary work, they were considered to be appropriate. However, during the detailed evaluation phase where only two sites were studied, questions were raised about the appropriateness of these records. Using the selection criteria discussed above, it was discovered that the Lake Hughes Array Station #12 was not located on bedrock. Instead, this station was situated above a thin veneer of alluvium. The character of the earthquake signal recorded at the surface was altered because it had travelled through the alluvium. Consequently, this record was excluded from the detailed evaluation phase. T A B L E 4.  Preliminary Assessment Earthquake Records Used SITE  LOCATION  BEDROCK  A(%g)  V(m/s)  A/V  Griffith Park  grd. level  granite  0.18  0.21  0.88  Cal. Tech.  basement  granite  0.19  0.12  1.66  Lake Hughes #4  grd. level  weathered granite  0.17  0.06  3.02  Lake Hughes #12  grd. level  thin alluvium  0.35  0.15  2.33  Note: All of these earthquake records are from the 1971 San Fernando Earthquake In the detailed evaluation, it was decided to use thirty-six records which met the selection criteria.  Several records (refer to table 5) were chosen, in order to provide a statistical  assessment of the effect of different A/V ratios and different earthquake types. Records used in the evaluation were from the 1971 San Fernando (magnitude 6.4), 1989 Loma Prieta (magnitude 7.1) and 1985 Mexico City (magnitude 7.9) earthquakes. The San Fernando and Loma Prieta earthquake records were utilized because they were close to the desired magnitude 7 event. These two earthquakes, however, represent strike-slip earthquakes.  28  In the area under investigation for this report, it is also anticipated that a subduction earthquake may be responsible for generating strong ground motion. As a result, it was decided to include records from a subduction earthquake event. The Mexico City records represent such an event, but they are recorded very far from the epicentre.  It is anticipated that the epicentre for the  subduction earthquake modelled in this report would occur much closer than in the Mexico City event. Therefore, Mexico City records hold limited value. 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CL  V* w  22  2 2 55  Cu fiu Cu Cu Cu Cu 03 03 03 03  j*v  K  >v|  K  1& =2 S a)  k.  X3  a o  2 S  e e to to  gs  1 / 3  •  to a o  1 I  >>  S o  0  o it  o  •o 03  Q Cu Cu  O  a-  '2  II  CO «>  r» r-  c  U.  JS .s  X X 09 o o is X X "eL"eL| 00 e c S 03 3 e e "o '3 9 9o k.0 0 k.0 0, ca u o o X X u S3 EL. C . at '£ I ) E E  '3 '3 X X  h. w Q  U,  — i  Ov Ov  Ov Ov Ov ov Ov Ov o< Ov  e  -o —  •vivo vd  -» •vi- •v* -» vo vo vo vd vd vd  ov ©N  X X  3  LU Q  z o <  o o  U  CO to  i-  ti  Du u u to CO  a a Cu-_ CL5a 3 < o U.  W  E  -O  00 X)  «  co iei -3  a  to  oo oo:  fi o o SS Ssl ka  T3  2 8 ^ 8 8 O T3  '  a o  to  *<o 2 O  <D  S i  il ii  cd x: H o C  O  30  CHAPTER 4 S O n , PROPERTIES INTRODUCTION Critical to performing an uncoupled soil seismic assessment, is the determination of the necessary soil properties which define the soil's pre-liquefaction behaviour, liquefaction resistance, and post liquefaction behaviour.  The soil properties, as determined by the soil's  behaviour in each of these three phases, form the basis of an uncoupled soil seismic assessment. The determination of the necessary soil behaviour information for each phase is independent, but in the solution of the seismic assessment the outcome of each phase is interrelated.  To  evaluate the soil's pre-liquefaction behaviour, the necessary information required for the dynamic assessment includes the maximum shear modulus (G  max  ), initial damping values (X), modulus and  damping reduction curves, which define the stress-strain curves.  To evaluate the liquefaction  potential, the soil's cyclic resistance is required and is based on the soil's density, static bias, confining stress and fines content.  The post liquefaction response depends on the post  liquefaction stress-strain and residual strength of the soil. The necessary soil properties for the performance of an uncoupled seismic assessment can be obtained by direct or indirect methods. undisturbed samples. used.  Direct methods refer to laboratory tests on  In the laboratory, such tests as resonant column and cyclic triaxial are  Indirect methods typically refer to in-situ tests which include SPT (standard penetration  test), BPT (Becker penetration test), CPT (cone penetration test), S C P T U (seismic cone penetration test with pore pressure measurements), and pressuremeter tests.  Of these two  methods, direct methods are limited in their use as a result of the extreme difficultly in and  31 associated high cost of obtaining undisturbed samples. Hence, to characterize the soil, indirect methods are more commonly employed. In general, the most widely used indirect tests include SPT, BPT and SCPTU, while the use of the pressuremeter is growing. For this thesis SPT and SCPTU results and experience were employed to determine the required soil properties for the three phases involved in an uncoupled soil seismic assessment. However, due to scope limitations of this thesis, only brief comments pertaining to correlations obtained using SPT and SCPTU were made.  For detailed discussions, current information  regarding use of the SCPTU and pressuremeter should be consulted (Finn, W., 1993, Robertson, P., 1990, Robertson, P., and Campanella, R., 1989, Seed, H . , and De Alba, P., 1986, Briaud, J., 1986, Robertson, P., and Hughes, J., 1985, Byrne, P., and Atukorala, U . , 1983; and Robertson, P., 1982). PRE-LIQUEFACTION SOIL PROPERTIES The primary properties required for the evaluation of the pre-liquefaction phase include the maximum shear modulus (G ), initial damping value (X), modulus and damping curves, and max  stress-strain curves. To determine the maximum shear modulus (G ), SPT N values (indirect) and shear max  wave velocities (direct) were used. In order to utilize the SPT N values, the values were first standardized for energy and overburden. The first step in standardizing SPT N values was to correct for energy using equation 21 (Seed, H . , et. al., 1984). N  60  = N  field  * (% Driving System Energy / 60%)  eqn. 21  The next correction to the SPT N values was to account for the effects of overburden stress (Liao, S. and Whitman, R., 1986, and Skempton, A . , 1986).  This correction was  32 obtained by determining the overburden effective stress at the specified N value depth and correlating it with an appropriate C factor (refer to equation 22 or the normalized equation N  22 A). C where  N  = /l/cr '  eqn. 22  v  cr '  is the vertical effective stress  v  (Liao, S., and Whitman, R., 1986; and Skempton, A . , 1986)  C  N  = \/pa/a '  eqn. 22A  v  Then, with these standardized SPT (Ni) values, shear modulus values for the different 60  soil types were calculated. For sands the following equations were employed (Seed, H . , et. al., 1986): G  m a x  where  = 1000 * 20 *  (N^o ' 1  * (O '  1 2  3  psf  eqn. 23  mean effective stress (cr, + a + <r )/3  a ' m  3  2  or G  m a x  where  = 1000 * (K ) 2  * (O ' 1  max  ( K ^ = 20 * ( N , ) ( K x W = 15 * ( N , )  1/3 60 1/3 60  psf  2  eqn. 24  (Seed, H . , et. a l . , 1986) or (for the Lower Mainland - Byrne, 1994b)  Normalized versions of these preceding equations are found below (see equations 23A, and 24A). G  m a x  = 22 * 20 * (N )  G  m a x  = 22 * (K )  where  x  2  1/3 60  * Pa * (<r 7Pa)  eqn. 23A  1/2  m  * Pa * (<r 7Pa)  eqn. 24A  1/2  max  m  ( K ^ = 20 * (Nl)m (Seed, H . , et. a l . , 1986) or (K^Uax = 15 * ( N , ) ^ " (for the Lower Mainland - Byrne, 1994b) P atmospheric pressure (101.325 k P a or 2116.217 psf) m  3  a  Generally, the (K ) 2  max  for gravels are greater than the (K ) 2  max  for sands.  It has been  33 suggested that for relatively dense, well graded gravels a typical (K ) 2  max  ranges from 75 to 135  (Seed, H . , et. al., 1986). In addition, the shear wave velocities, obtained using SCPTU, were used to calculate shear wave modulus in a direct manner. The shear wave modulus was determined using the following relation: G where  v, p  m a x  = p *v  2 s  eqn. 25  shear wave velocity soil density  Studies have concentrated on the determination of shear modulus values for sand, gravel, clay and peat (Seed, H., and Idriss, I., 1970, and Seed, H., et. al., 1986). Additional research has been conducted to determine the modulus values of silts, organic silts and polystyrene (Byrne, P. et. al., 1984 and Byrne, P., and Atukorala, U . , 1985). The following table (refer to table 6) provides a brief summary of the available information. Damping values for the various soil types encountered can be obtained in-situ by interpreting the unload-reload curves generated during a pressuremeter test, but for this thesis the damping values were estimated (refer to table 7).  34  TABLE 6.  Table of Proposed Shear Modulus Values for Various Soil Types (from * Byrne. P.. et. al.. 1984. and " Byrne. P.. and Atukorala. 1985)  Depth (m) - (ft)  Shear Modulus (G ) - kips/ft  Shear Modulus (G ) - MPa  Peat*  1.8 - 2.4 m 6 - 8 ft  25  1.2  Peat*  1.8 - 2.4 m 6 - 8 ft  17  0.8  Organic Silt*  7.4 - 7.9 m 24 - 26 ft  86  4.1  Organic Silt*  10.4 - 11.0 m 34 - 36 ft  194  9.3  Hog-Fuel*  -  112  5.4  Polystyrene"  1 - 5.5 m 3 - 18 ft  174  8.3  Peat**  5.5 - 9.5 m 18 - 31 ft  26  1.3  Organic Silt**  9.5 - 21.5 m 31 - 71 ft  87 - 191  4.2 - 9.2  H a r d Clay**  85 - 106 m 279 - 348 ft  5555  266  Till**  106 - 246 m 279 - 805 ft  20885  1000  Soil Type  2  max  max  35 T A B L E 7.  Summary of Available Recommended Maximum Damping Ratios (X ) for m3X  Various Soils (from ' Seed, H., and Idriss, I.. 1970, " McCammon, N.. et. al.. 1990, * Seed, H.. et. al., 1986. Byrne. P., et. al.. 1984. Byrne. P.. and Atukorala. 1985) g  9  Material Type  Ratio (X )  Clays*  26 - 32 %  Sands*  21 - 28 %  Sands**  28 - 30 %  max  Gravels, gravelly sands ''  26 - 32 %  Peat  27 %  1  1  Organic Silt Hog-Fuel  5  5  27 % 53 %  Polystyrene'  13 %  Peat'  27 %  Organic Silt'  27 %  Sand'  25 %  Clay-Silt'  15 %  Dense Sand-Gravel'  13 %  Hard C l a y '  13 %  T i l l L i k e Soils'  11 %  Extensive research has been conducted in order to develop models which attempt to predict the reduction in the soil's shear modulus and increase in damping values during seismic loading and corresponding soil straining. Researchers (Hardin, B., and Drnevich, V . , 1972, and Seed, et. al., 1986) have developed shear modulus reduction and damping increase curves. The formulation of these curves has been based upon laboratory testing which has included resonant column, cyclic triaxial, and simple shear tests. The modulus reduction curves are based on the mathematical expression (Hardin, B., and Drnevich, V . , 1972):  36 G/G where  7  m a x  = 1 /(l+  7 h  ).  eqn. 26  hyperbolic strain  h  In addition, the damping curves for the various materials were determined to be described by the following equation (Hardin, B . , and Drnevich, V . , 1972): / -Vax = 1 - G / G where  X X  m a x  m a x  .  eqn. 27  fraction of critical damping maximum value o f X that occurs at large strains  For this thesis, the shear modulus reduction and increasing damping curves used were the recommended curves which were a part of the S H A K E program. LIQUEFACTION  RESISTANCE  For this thesis, the determination of the soil's liquefaction resistance was based on in-situ SPT and S C P T U results together with correlations with such factors as density, gradation, fines content, plasticity, confining stress and static bias. It should also be recognized that the soil's liquefaction resistance can also be obtained from BPT (Becker penetration test) and laboratory testing of undisturbed samples (Finn, W . D . , 1993).  As discussed previously, obtaining  undisturbed samples is extremely difficult and is not usually considered practical. SPT N values were obtained at all of the sites, and were corrected for energy and overburden stress, yielding ( N ^ .  These ( N ^ values were used to determine the cyclic  resistance (CRRJ of the soil which is a total stress method of defining liquefaction resistance. Discussion of how the cyclic resistance was then used in liquefaction assessment was provided in chapter 2. SCPTU records were only obtained from the Tsawwassen and Ladner sites.  These  records were interpreted and correlated with available information to generat ( N ^ values which  37 were then used to calculate a cyclic resistance ratio. POST LIQUEFACTION SOIL BEHAVIOUR After it has been determined that a soil layer liquefies, it is accepted that the strength and stiffness of the soil is significantly reduced and the stress-strain curves are significantly different than before liquefaction has been triggered. Therefore, the first step in the post liquefaction analysis is to determine the reduced strength of the liquefied soil layer. Calculating the strength of the liquefied layer can be based on direct methods such as laboratory testing (Finn, W., 1993, Salgado, F., and Pillai, V . , 1993, Poulos, S., et. al., 1985) and indirect methods such as the correlation of historical evidence with in-situ field tests (Seed, R., and Harder, L . , 1990, and Seed, H . , 1987). Laboratory test results have indicated that a strong interdependence between the residual strength S  us  and the effective overburden pressure a'  vo  exists. Results have yielded that the  residual strength may calculated using equations 28 and 29, when ( N ^ values are approximately 4 or equation 30, when ( N ^ values are approximately 13. S  us  = 0.075 a\  eqn. 28  Q  (with a minimum value of 5 kPa - Finn, W., 1993) S  us  = 0.087 <r'  vo  eqn. 29  (Byrne, P., 1990c) S  us  = 0.21 a\  Q  eqn. 30  (Salgado, F., and Pillai, V . , 1993) In summary, the residual strength is proposed to be a function of preliquefaction soil density and effective stress and can be simply determined by equation 33.  38 Sus  —  a  where  S  m  a a\„  T A B L E 8.  eqn. 31  vo  a  post liquefaction residual strength at limiting strain strength coefficient which is a function of density (see table 8) vertical effective stress  Strength Coefficient Table Based on ( N ^ Values and Limiting Strains (Byrne. P.. 1994b)  a  7LIM%  0 -4  0.07  100  4 - 13  0.07 - 0.2  100 - 20  > 15  0.6  < 20  An alternate approach to determine the residual strength is based on a review of a number of case histories where large deformations occurred after liquefaction had been triggered (Seed, R., and Harder, L . , 1990). Based on this review, a correlation was developed between the insitu (NJgo values and the back calculated residual strengths. Interestingly, it was observed that an increase in fines content for the same ( N ^ value would result in a larger residual strength. Therefore, the (N ) values were corrected for silt content (refer to table 9 and equation 32). x  60  (Nl)60 effective  TABLE  9.  =  (N^o measured "•"  A  (Nl)eo  e  Correction ( A N ^ Values Due to the Presence of Silt L.F.. 1990. pp. 370)  Fines Content (%)  (AN^O  10  1  25  2  50  4  75  5  Q - 32 n  (from Seed. R . B . . and Harder.  39 This empirical prediction method was based on cases where the confining stresses ranged from 50 to 150 kPa. However, for larger confining stresses, the empirically based method does not agree with laboratory results (Byrne, P., et. al., 1993). Thus, for confining stresses greater than 100 kPa where historical evidence is unavailable, the laboratory method appears to provide acceptable results. After determining the residual strength of the post liquefaction soil, the next step is to ascertain the post liquefaction stress-strain curves. As discussed in chapter 2, two different stress-strain curves have been employed. One method modelled the soil's stress strain behaviour with a rigid plastic spring which does not accurately model the soil stress strain response. As a result, a modified stress-strain curve was proposed (Byrne, P., 1990c) which modelled the stiffness and residual strength of the liquefied layer with a nonlinear spring (refer to figure 6). With this modification, it is recognized that at the moment liquefaction is triggered a dramatic reduction in stiffness occurs. As a result, the liquefied soil is softened, allowing the static driving stresses to move the soil. When the soil moves, it will have an associated velocity and consequently, will move an additional amount due to these dynamic forces.  The entire  movement, however, will be subject to a balancing of the work done. Comparing stress strain curves, it should be noted that the simplified stress-strain curve neglects the displacements from P to S (Jitno, H., and Byrne, P., 1994). Strains generated between P and S could range from 20 to 50 %, depending upon the soil density and consequently, displacements could be considerable before the soil's residual strength is achieved. For this thesis both curves were utilized and compared.  40  CHAPTER 5 SITE LOCATION. TOPOGRAPHY. VEGETATION. AND GEOLOGY INTRODUCTION For this thesis, five Ministry of Transportation and Highways bridge sites were identified as sites requiring seismic assessment surveys.  These sites were the Steveston Underpass,  Blundell Underpass, Ladner Exchange, Mathews Road Underpass, and Tsawwassen Overpass. In this chapter, discussions pertaining to the specific site locations, and surfical geology at the sites are presented. SITE LOCATIONS All  five sites are located in the Lower Mainland region of British Columbia.  Specifically, the sites are situated within the municipal boundaries of Richmond and Delta, near the western edge of the Fraser Delta (refer to figure 7) and are located along a major north south tranportation corridor. SITE LOCATIONS - Steveston Underpass The Steveston Underpass is located where Steveston Highway traverses over Highway #99 (refer to figure 8).  The underpass is located in the municipality of Richmond and is  approximately 1 kilometre north of the north portal of the George Massey Tunnel. SITE LOCATIONS - Blundell Underpass The Blundell Underpass is located where Blundell Road crosses over Highway #99 (refer to figure 8). The underpass is located in the municipality of Richmond and is approximately 3.5 kilometres north of the north portal of the George Massey Tunnel.  41  Figure 7 Area Plan - MoTH Site Locations Scale 1:250,000, Map 92 G (Surveys and Mapping Branch, Department of Energy, Mines and Resources)  42  Figure 8 Location Plan - MoTH Site Locations Scale reduced from 1:50,000, Map 92 G 2/3 (Surveys and Mapping Branch, Department of Energy, Mines and Resources)  43 SITE LOCATIONS - Ladner Exchange The Ladner Exchange consists of Highway #17 traversing over Highway #99 (refer to figure 8).  The underpass is located in the municipality of Delta and is approximately 1.5  kilometres south of the south portal of the George Massey Tunnel. SITE LOCATIONS - Mathews Road Underpass The Mathews Road Underpass is located where Highway #10 traverses over Highway #99 (refer to figure 8), in the vicinity of the southern side of Burns Bog. The overpass is located in the municipality of Delta and is approximately 9.5 kilometres south west of the centre of North Delta. SITE LOCATIONS - Tsawwassen Overpass The Tsawwassen Overpass is located on Highway #17 and crosses over the British Columbia Rail (BC Rail) line and Deltaport Way (refer to figure 8). The overpass is located in the municipality of Delta and is approximately 4.5 kilometres north east of the town centre of Tsawwassen. GROUND TOPOGRAPHY AND VEGETATION Topographic conditions and vegetation at all sites are similar.  The ground surface  surrounding the bridge embankments is level. Typically, variations in relief are less than 1 metre and are related to road grades being elevated above the natural ground surface or shallow drainage ditches. At the Blundell Underpass, ditches 2 to 3 metres in depth are identified on both sides of Highway #99 but they terminates within a few metres of the embankment. The ground surrounding all the sites, except the Blundell Underpass, is vegetated with grass and the occasional small shrub. At the Blundell Underpass, the area surrounding the  44 embankment is thickly vegetated with mature trees and shrubs. GROUND WATER CONDITIONS Available drill records and cone penetrometer test results indicate that the ground water levels at the time of the investigations were within 1 to 3 metres of the natural ground surface. Ground water levels are known to change with climatic and seasonal variations. Typically, the water level is within 1 metre of the surface. As a result, a conservative approach was used in the analysis by assuming that the water table at the time of the earthquake would be close to or at the natural ground surface. SURFICIAL GEOLOGY - General These sites are situated within the western portion of the Fraser River Delta and consequently, are comprised of Fraser Delta sediments. The surficial soils which comprise these sediments include fluviated, interdistributary swamp, salt marsh, and shallow water deltaic and deep water pro-delta facies, overlying Pleistocene deposits (Blunden, R., 1975). These deposits are late pleistocene and are locally inset into and on top of the earlier soils, Vashon tills and bedrock of the Oligocene and Miocene age. The delta development has been influenced by isostatic, eustatic and tectonic forces that have existed in this area.  These forces were  responsible for causing this delta to develop and become emergent approximately 8,000 years ago. The surficial deposits in the Richmond area consist of thin discontinuous veneers of clays, silts, and peats which were deposited in fluviatile, interdistributary swamps, and salt-marsh environments (Blunden, R., 1975). These soils overlie a thick sequence of sands and silts which have reached an approximate maximum thickness of 30 m (100 ft). The exposed deposits were  45 formed in the intermediate and lower tidal flat zone. These deposits are underlain by a pro-delta series which was deposited in deep water and is bedded against the proto-delta. These marine delta deposits reach a maximum depth in excess of 210 m (700 ft). A recent deep drill hole and reflection seismic surveys have provided excellent detailed information concerning the subsurface stratigraphy (Luternauer, J . , 1991b). Specifically, at the Blundell Underpass, Steveston Underpass, Ladner Exchange, and Tsawwassen Overpass sites, the deposits include deltaic and distributary channel fill sediments overlying and cutting estuarine sediments and overlain by overbank sediments (Geological Survey of Canada - Map 1486A, 1979). The overbank sediments consist of silt to silty clay loam normally less than 2 metres thick overlying 15 metres or more of interbedded fine medium grained sand which may contain minor silt (Geological Survey of Canada - Map 1486A, 1979). Salish River sediments are underlying the Mathews Road Underpass and consist of bog, swamp, and shallow lake deposits (Geological Survey of Canada - Map 1484A, 1980). Specifically, these subsurface soils are believed to consist of lowland organic sandy loam to clay loam, 0.15 to 0.45 metre in thickness overlying medium to coarse sand (Geological Survey of Canada - Map 1484A, 1980). Also shown on the surficial geology map is the presence of a small patch of Fraser River sediments which may underlay the northern embankment. SURFICAL GEOLOGY - Steveston Underpass - Soil Conditions A general soil profile at the Steveston site consisted of a surface layer of topsoil or peat (average thickness of 0.6 metres), underlain by silt (average thickness of 3 metres). The silt is overlying a fine to medium grained sand which includes some silt lenses. Borehole information, available from MOTH files, was utilized to construct a preliminary  46 subsurface profile for this site (refer to table 10). The depth to firm ground was estimated to be approximately 30 meters (100 feet). At the time of the site visit on June 29 , 1991, topsoil was the only exposed soil, and water levels were within approximately one metre of the ground surface. SURFICAL GEOLOGY - Blundell Underpass - Soil Conditions A general soil profile for the site would consist of a surface layer of topsoil or peat (average thickness of 0.6 metres), underlain by silt (average thickness of 3 metres). The silt overlays a fine to medium grained sand which includes some silt lenses. For this site, MOTH drill records were used to develop an preliminary soil profile (refer to table 11) for the Blundell Underpass site. Similar to the Steveston site, the distance below the ground surface to firm ground was estimated to be approximately 30 meters (100 feet). At the time of the site visit on June 29, 1991, soils exposed on the surface and within the ditches appeared to be a mixture of silty sand and topsoil. Ground water levels were within approximately one metre of the ground surface. 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CO  5  r-  3  rt  I  E  £  co  s u  IO  rr  CB  CO  8 5to sr ro*~— rrpco & 3| w  CS u CO  to  CO  ft,  CM CO  CB  'S co  S  o  ^  o n io in" o » o «" io sr o S" r» O CD CM C O C* ~M C O CM C B C M C O . <r* "~ «™  o « io »-  CO  N™  r-  V -  Ul  o  ^2  LO CM  oo  63" o rT to co CM CB — rC B CO o . <  §f *~  CM_  to  CM"  CO  » .  *~  CM  o rr" CO o  ft(m)  Thickness ft(m)  z 1  Depth  8  rr CB  r-  rr CB  co to  rr  IO  to  <o  to r-  CM CO  CO  o  E  „  to-  " si  rf  CM CM  I"  .  co «. rr o.  o  00  52  N  rr « rco CO C M r» C M tO 1 1 1 o I I I ' 1 1 O 1 «- o 2 " S r» co C M rr o 2 o  s  CM  co  rr  10  CM" CM  co  CO  1 CM  co"  o  CM  «, s CM  oo. o  rr  CO  1  TT  >  1  1  CO" T-  1  T-  IO'  IO  r-  1  C*j S 3 3 C O 3 b ad  1  ; CO *—  >n to  tc o  if  p CO  CM  •o  st 1  1  * -  IO  IO  IO  oo  co" r- •-. rf  8 oi •-•r?  CO  to *- 10 1 1 1 C O hCM io O r-' to  IO  co  00  CB  o  <0  5  CM  CM  1  O Co  _,  to  to  E B  to p  CM CO  r» 1  CB CO  CM CM  1  rf  O  r» CS  CO  CM  CO*  CO  CM  1  1  1  CO  CO  CO  r~  CM  r»  CS CN  CM  CO*  CO CM  rr  otT to  00  CO  ^ rr  <D  O  CM  CM CM  JM,  1  rN -  co"  CO CM  1a ki  CO CM,  CM  IO  CO  -o  O a o  01  CQ  s1  CB  s  « »  e  IS O  03  s  Layer  s  CM CM  CO  03  £M.  Cu.  §~  £3  w oa <  49 to be close to the existing ground surface. During the fall of 1991, a SCPTU hole was advanced near the north east corner of the underpass. Based on an interpretation of the SCPTU information, a soil profile was developed (refer to Appendix A). It was determined that the soils within approximately 1.5 metres of the surface consisted of a sandy silt and sand mixture. This layer was then underlain by a clay layer which was approximately 0.5 metres thick. The clay was underlain by a variety of layers which were composed of sandy silt, silty sand and sand. The density of these cohesionless soils was interpretated to be loose to loose/compact.  Pore pressure information indicated that the water  level was located approximately 2 metres below the ground surface. The soil types identified with the SCPTU generally agreed well with the soil conditions identified by the drill logs and SPT information. A combination of MOTH and GSC information was used to derive a preliminary soil profile (refer to table 12). Using this information, it was hypothesised that the firm ground may be situated at a depth of approximately 90 meters (300 feet). To determine a detailed soil profile, one SCPTU hole was advanced to a depth of 50 metres. Using the CPTINT program, the cone data was interpreted and soil parameters were created for the near surface (refer to table 13). For soil information beyond the depth penetrated by the cone, seismic reflection survey and deep drill hole information was utilized. Values of the soil parameters employed in the densified zones are also included in the table. SURFICAL GEOLOGY - Mathews Road Underpass - Soil Conditions The subsurface soils consisted generally of 0.6 to 3.0 metres of peat, then 1.0 to 3.4 metres of soft organic silt which was then underlain by a mixture of sand, and sand with silty  ? YT 7  -1 2  3  SM  i  SM  2  CO  io  ?  7 f  7 ?7 f ? f 7 ?77 7 f ZZ.  o  IO IO  8  IO IO  8  o  to  8  io  t-  8  IO IO  IO IO  IO  IO  CM  IO  IO  IO  IO  IO  IO  IO  IO  to  IO  d  d  d  d  d  IO  d  IO  d  IO  d  CO  8  CM CO  CM  8  CM CO  CM CO  8  o  o  CM  a S, s.  i£ 8  CM  co  £ ST  i o co" 8  a  _i  2 CO  at a> co SjT to «o UI UI UI UI UI UI UI UI UI UI UI UI UI UI UI UI UI UI UI UI UI UI © o o III UI UI UI o o o o o o o o o o o o o o o o o o o o o o o o o CO' 3 , co" »£, co' j £ . CO' j £ co" £ CO CO' j f . CO' co j f , co" j £ CO* j f CO jf. co" co"  at  0 |  CO  to  d  Ss  CO  SP/SM  CO  z  a.  i  DL  SM/SP  SOIL K ft/s(m/s) TYPE 05  Q  _t 2  11 CM  3.  IO ! ? CM „ • CM  s s  03  8  f». 2  d  d  CM  co  d  d  d  o  3  CO  CO  r-  co  tv ®" CM CO 9 ^ co co* i o S* co •» jo co" io a> co' o ? 9 co & 2 CO v~ CO JM,  5  8d  8  CM CO  £8  8  CM IO  CO  ?E 8  to  cn  E  ca Z  £>2  o  ^ o O  tC io  "> 5 °> Z  u  o O  R  o  io  « r- o  m  IO  sr  a "**•»  O CO i o oo •*_• r - O  io j ;  PT  «T  o  CO  O  O CO IO  -  to  ^ o co* CM cO  2  *  V*  CL,  o>  Thickness ft(m) Depth ft(m)  to  CM"  w  CM"  to « ; 1 "O 1 o  „  O  2? o  O  o O  *" a" S £ 2 •*  CM  l  IO  o  CO CM  CO  CM  -  o o CO "  CM  FT  rl  a2  *°.  g 2.  S" • * 2.  5T  CM"  rco i o co'  SCM  •o *-  8  CO CO  2 2  co"  o*  CM" — CM"  c\  S  •  2 i  88  „ o  o>  CO  io _ o O  07  CO* CO  CO*  o  CM  8  cf 00  CO CO* CO CM co  1  3  co'  at o CM  d § si  1 3 CM 5 I I » CO CO  d oo da  Layer  CO  IO  co  CO  a  at  CO T - CO  1  jjo*  8 CO*  ^ 1 1 o o o>  1  —  ,  CO CO CO  o cn "~ CM CM IO CM  CO  JM,  CM  s* _ s* °d  3"  O CM •0- CM  CM  c l " si " si *~ si  to  CM  i 1 1 i i n r>- IO CM  t i  IO  CM CO  CM  -  1  a 1 1 o CO  8£  " d •CM  ». o  8 8 81  JO  O  CO*  CO*  o  CO  co'  5  a  ~8  1  o  ^  5  3  •r, <N  vo o v <N  CA  Si  ss:::*!*:  o rm t—(  •n VD <N  ov  O  v~ Tf  i l l  .vpix-NO'  00  roi  VD <N  ov  52 layers which varied in thickness. Both available MOTH and GSC information was employed to create a likely soil profile (refer to table 14). Also, as at the Ladner Exchange, the firm ground was estimated to be at approximately 90 meters (300 feet) below the ground surface. At the time of the site visit, topsoil was observed in drainage ditches situated approximately 50 metres from the bridge fills. Based on information provided in the foundation report, the water table at the time of the original site investigation was observed to be at a depth of 0.5 to 2.0 metres below the ground surface. SURFICIAL GEOLOGY - Tsawwassen Overpass - Soil Conditions For this site, a generalized soil profile includes a near surface horizon of loose silty sand (6.4 metres thick), underlain by a medium density layer of sand (2 to 3 metres thick), which is in turn underlain by a sequence of loose sand with some silty layers and silty sand, and is then underlain by medium dense sand. A combination of MOTH and GSC information was utilized to construct a preliminary soil profile for both the north and south ends of the Tsawwassen Overpass (refer to tables 15 and 16). Detailed soil information for approximately the first 200 feet of the profile was obtained from the MOTH drill logs. However, determining the depth of the firm ground layer involved extrapolating the deep drill hole information (Lutenarer, J., 1991b) with the available reflection seismic survey profiles.  Based on this information, it was estimated that the depth to firm  ground, where the base acceleration would be applied, was approximately 200 meters (600 feet). These soil profiles, however, were only used in the preliminary assessment. During the fall of 1991, a SCPTU hole was advanced near the north east corner of the  -  CM  co  "Of  to  CD  co CO  to"  CO  (2.44)  a  o  CM  CO  6  rr  to  CO  150 (45.73)  300 - 600 (137.20-182.93)  rr  r-  CO  11050  8  a  (529.1)  CO  130  0.45  o  (20.4)  S  8604  8  (412.0)  o r-  125  E-9)  E-9) (1.0 E-8)  3.0 E-8  (1.0  3.0 E-9  LU LU  (19.6)  (1.0 o o to j £  150  E-8 E-9) 3.0 E-9  (5.0  1.6  (5.0 E-8)  1.8 E-7  E-9)  3.0 E-9 (1.0  Tf  (45.73)  0.45  3.0 E-9 (1.0 E-9)  ML  MUSM  SP/ML  ML  ML  ML  CO  300-450  8  5686  3.0 E-4 (1.0 E-4)  2'  (91.48-137.20)  8  (272.3)  CO CM  120  b  (18.8)  IO  128  2685 (128.6)  0.45  o  (39.02)  CM CD  120  CM CO  (18.8)  4108 (196.7)  0.45  "Of  172 - 300  115 (18.1)  3974  IO  (52.44-91.48)  22 (6.71)  150-172 (45.73-52.44)  8  10  to  (3.05)  CM  140-150  CM  o  (190.3)  0.45  o r»  (42.68 - 45.73)  rr  115  IO  (18.1)  3231 (164.7)  CO  (1.83)  134-140  2517  0.45  o r>.  (40.85 - 42.68)  3 CM  115  to rr  (18.1)  CD  24  r~  (120.5)  o>  (7.32)  CM  CM  110-134  to  8  (33.54 - 40.85)  IO  110  CM CO  (17.3)  to  10  to  (3.05)  IO  100-110  to  3367  o rr  (181.2)  CO  b  120  o o co'  (18.8)  SM  a. (0  (30.49 - 33.54)  co  f  16  CO  Ul Ul  (4.88)  (1.0 E-6)  SM  CO  84-100  to 2605  to  (124.7)  b  T  (17.3)  2050 (98.2)  3.0 E-6  (1.0 E-6) 3.0 E-6  SM  2  (25.61 -30.49)  *~ r-  to  78-84  o to  (23.17 - 25.61)  rr rr  105  8  (16.5) 110  b  24  IO  (7.32) 8  8  52-76  s 1537  co  (73.6)  CM  100  j£.  (16.85-23.17)  CM  «? f  (15.7)  rr  o o co'  4  CO  LU UJ  (1.22)  0.45  3.0 E-6 (1.0 E-6)  CO  48-52  co  2098  o r»  (100.5)  3.0 E-6 (1.0 E-6)  SM  2  (14.63-15.85)  o  CO  0.5  o  42-48  rr a> 1401  s  (67.1)  b  100  Tf  (15.7)  8  (12.80-14.63)  •or  4  to 2659  3.0 E-6 (1.0 E-6)  CO  (1.22)  o (127.3)  E-8 E-9) 3.0 E-9 (1.0 E-9) 3.0 E-9 (1.0 E-9) 3.0  (1.0  2  (11.58-12.80)  r-  120  ~ fc.  (18.8)  £M  38-42  4  3  (1.22)  34-38  0.45 CO  o  (10.37-11.58)  5 110  IO CM  E  (17.3)  rf  16  300 (14.4)  Q  (4.88)  25 (1.2)  05  18-34  25 (1.2)  K SOIL ft/s(m/s) TYPE  o  (5.49-10.37)  70 (11.0)  8  (3.05-5.49)  CO "Cf CM,  10-18  8  CM  (2.44)  u>  (0.61 -3.05)  (11.0)  70  E ca Z  2-10  2  rf  (0.81)  0-2  1  (0-0.61)  Thickness ft(m)  Depth ft(m)  A  £  SO  Layer  53  2  CO  to CO  es a,  vV TJ  rr  b  TJ  e  CC  O  03  CO CO  0 o  * IT  (Q Si  CO  r«.  u c  8  1I  —  3  E  co W  CQ <  — )  S  o cu g co  <« J3  . tS  E 2  cS  O ° CO T J  £  o  _ u  £ «  TJ to «« oaCu <*  o „  | w  Layer  §~  5,1,  •iv  CM  8 <s CM  8  s  8 5  *—  CM  co o» co  u  o  o a  *~  <0>  0.45 0.45  CM  IO  CO  3  IV IO  CO  IO  3  CM  IV  3  CO  n  CM CO 1  co' O  a  s  TT CM IV CM  $  1  SM  E-6 E-6) E-6 E-6) E-5 E-5) E-5 E-5)  o o co' V  3 o  to  00  CM  o  CM  co  3 CO  8 co  3 CM  co  8  |v  -  CO  m  IO  CM  CM  o m  T CO  >o>  CM  3 3  co  IO CO  in  CO  3  8  IO rv  3  to  SP  SP  SP  SP  SP/SM  SM  SP/SM  SP  SP  SP  SP  CO CO UI UJ  3.0 (1.0 3.0 (1.0 3.0 (1.0 3.0 (1.0  8 0.45  o  0.45  to  0.45  0.45  3 0.45  8 •flea  0.45  O  0.45  IO IV  115 (18.1) 110 (17.3) 100 (15.7) 115 (18.1) 120 (18.8) 115 (18.1) 110 (17.3) 125 (19.6)  o  SP  SP  SM  E-6 E-6) E-5 E-5) E-4 E-4) E-4 E-4) E-4 E-4) E-4 E-4) E-5 E-5) E-6 E-6)  3.0 (1.0 3.0 (1.0 3.0 (1.0 3.0 (1.0 3.0 (1.0 3.0 (1.0 3.0 (1.0 3.0 (1.0  o j£.  53  IO CM  0.45  SM  cj  0.45  8  0.45 o  105 (16.5) 110 (17.3) 110 (17.3) 120 (18.8) 115 (18.1)  585 (28.1) 759 (36.4) 1242 (59.6) 1798 (86.3) 1891 (90.7) 2981 (143.0) 2381 (114.2) 2261 (108.5) 2793 (134.0) 2484 (119.2) 1905 (91.4) 3162 (151.7) 4042 (193.9) 3402 (163.2) 4990 (239.4) 9076 (435.4)  SOIL TYPE  K ft/s(m/s) f ?  74-78 (22.56 - 23.78) 78-84 (23.78 - 25.61) 84-118 (25.61 - 35.98) 118-122 (35.98 - 37.20) 122-132 (37.20 - 40.24) 132 - 200 (40.24 - 60.98) 200 - 600 (60.98-182.93)  co Z 100 (15.7)  IT  4 (1.22) 6 (1.83) 4 (1.22) 20 (6.10) 6 (1.83) 4 (1.22) 4 (1.22) 12 (3.66) 4 (1.22) 6 (1.83) 34 (10.37) 4 (122) 10 (3.05) 68 (20.73) 400 (121.95)  Thickness ft(m) 05 Q  0-14 (0-4.27) 14-18 (4.27 - 5.49) 18-24 (5.49 - 7.32) 24-28 (7.32 - 8.54) 28-48 (8.54 -14.63) 48-54 (14.63-16.46) 54-58 (16.46-17.68) 58-62 (17.68-18.90)  Depth ft(m) UI UI  UI UI UI UI  Y f f ?  o o o o co' j£ CO  ZZ.  in  CO  d  d  o  09  CO |v  CO  a  co  r  Laj er  Q« 0.45  8  rr _ rr s  CM IO CO  rr  CB CM  rr rr  CO  CO  3 CO  CO CM  rT  rr  •  CO  co  CO  o  CO  CM CM  • CO  CM CM C O  1  rr  oo CO  CO  So  CO*  ti ss 1  0.45  8 d 0.45  rr  d 0.45  o  0.45  B  0.45  rr  CO  CO  r»  CB  rr  CM CM  3  r%3 co8  d  CO  IO CM  to rr  r» rr  CO  co co CO  rr  rr CO r r y-  r— rr  8  3 CO CM  8 CO CM CB CM  r~  IO CO  CM  co  r»  CM CM  rr t rr r r  CO  rr  oo CM CM  ss  rr  8 m r»  rr  S  co  CM  '-co-  S  | |  CM  in co IO  o rr o rr m m  CO  CO  rr rr  d d d  3 8 CB CM  00  o r»  co  rr rv i  S  IO°  o  s  CM  rr o  s  CM CM C| O  CB C O CB* CB CM  •  CM  CM  CB  rr CO  CO  CM IO  r»  _ in  O ©  co co oo. r r  CO  rr  3  CM CO  o  CO  8  to  CO  CM CM  o <n  8  m  - ~ CO  *z  CO  » N co oo co  IO  rr  09  S 8  CM CO  o  CO  in  co CM  CM CM  rr  r~ CM CM  8 o  CO  00  0.45  0.45  O  0.45  co"  0.45  O  0.45  TT  0.45  L U U I  zz.  3.0 E-4 [1.0 E-4) 3.0 E-4 (1.0 E-4) 3.0 E-4 (1.0 E-4) 3.0 E-4 (1.0 E-4) 3.0 E-5 (1.0 E-5) 1.6 E-6 (S.O E-7) 3.0 E-5 (1.0 E-5) 3.0 E-4 (1.0 E-4) 3.0 E-4 (1.0 E-4) 3.0 E-4 (1.0 E-4) 3.0 E-5 (1.0 E-5) 1.6 E-6 (5.0 E-7) 1.6 E-6 (5.0 E-7) 3.0 E-8 (1.0 E-8) CM CO  co  00  t  CB  SM  SM  >M/SP  SP  SP  SP  SP  SP  SM  SP  SP  SP  SP  SP  SP  SP  SP  SP  3.0 E-5 [1.0 E-5) 3.0 E-4 [1.0 E-4) 3.0 E-4 [1.0 E-4)  3.0 E-6 [1.0 E-6)  :/s(m/s)  K  o oV c o .  s 5.  is.  400.  CO CO  rr rr  d  CO  CO  or  i  o  CM  (121.98'  s L U U I  68 (20.73) 200 (60.98) 200 (60.98)  o  0.  120 (18.8) 115 (18.1) 120 (18.8 115 (18.1) 120 (18.8)  CO CM  f f  74--88 (22.56-- 26.83) 88(26.8398- 104 (29.88 --31.71 104--108 (31.71 •- 32.93) 108--124 (32.93 •- 37.8( 124--128 J7.80 •- 39.02) 128--132 (39.02 --40.2' 132--200 10.24 --60.91 -400 121.8  w 346 (16.6) 793 (38.0) 2386 (114.3) 1711 (81.9) 2244 (107.5) 1573 (75.3) 2043 (97.8) 1425 (68.2) 2030 (97.2) 3158 (151.2) 2686 (128.6) 2931 (140.4) 2119 (101.5) 3764 (180.2) 3005 (143.9) 3781 (181.1) 3274 (156.8) 4600 (220.3) 6980 (334.2) 9371 (448.7)  ° i cu  90 (14.1) 95 (14.9) 120 (18.8) 115 (18.1) 120 (18.8) 110 (17.3) 115 (18.1) 100 (15.7) 110 (17.3) 120 (18.8) 115 (18.1) 110 (17.3) 95 (14.9) 120 (18.8)  8  a  10 (3.05) 14 (4.27)  pc  rf •a-  -48 -14.63) -52 -15.85)  ft(m)  Q  0-8 (0-2.43) 8- 24 [2.43-•7.32) -28 -8.54)  Thickness  o r co H 1  m  d  o rr  CQ  8  y~  O  82  s  CM  i  56 underpass (refer to Appendix B). Interpretation of the SCPTU records revealed that the soils within 4 metres of the surface were a clayey silt and sand mixture. Underlying this horizon, an assortment of silty sand and sand layers were identified. These cohesionless soil layers ranged in density from loose to loose/compact. Generally, these soil types identified with the SCPTU, agreed with the soil conditions identified by the drill logs and SPT information. At this site, deep soil information was obtained by advancing a SCPTU hole to a depth of 50 metres.  As discussed, readings obtained with the cone were interpreted and soil  parameters for the near surface were developed. As for the preliminary analyses, deep soil information was obtained using available reflection seismic survey profiles and deep drill hole information, which were available from the Geological Survey of Canada. Utilizing this information, a soil profile was created (refer to table 17) which was used in the detailed SHAKE analysis. In addition, soil parameters for the densified and embankment soil were estimated using engineering judgement and available information and guidance (Byrne, P., 1992). These parameters were also utilized in the simplified deformation analyses. For the complex deformation analyses, soil parameters were also estimated using the cone information, engineering judgement, available information (Duncan, J . , et. al., 1980), and my thesis advisor's guidance (Byrne, P., 1992). A summary of the soil parameters used in the complex deformation analyses can be found in table 18. Based on the borehole log information, the ground water levels at the time of the first site investigation were estimated to be at the existing ground surface.  57  Mi  8 <l  :jSS:0>:  - '"i lag*  : :  '-<£>::: *j;:  01  <?l  IP 1  60  2  <"M  VI W* i^'SSi  <5-  >&i in  yoi: ol MM:  ^  V)  ^  Ol ol  ol  cK  5?  oi  <S  oi.  o  cS  ^  i| ii  (51 :  : : : : :2rj :  ;  ;  m  o  ol  ;  Co  1  IQ  «l  —i  o o  *  0\  fi  o  O  o  O  s&KiKis;  -  ^  CO  5  I2  m  5  * 2  ;;;*;;;  o »n  >n oo  Ov  :;:#::::  d  ;;j;;;;;;;  5  O «n  =  d  ~  O  in  ON  JN  od  vo  II © I 5  o  <=  5  fN" S:;:;:;:;:;:;:;:;:  00  •ri  -H  00  s °° * m$.  ON  s  O  in  O  +  0\  9  m  : :  ( N o *  I  s  vn  CM  «n  SM/SP  SM  SM  O  ::;*::  S^SsSs:  ©  ©  o o  o O o  VO  § 1 1§ 1  % III  o  J?  I  SM/SP  »n  d  1  SP/SM  rt  * 2  40000  s  ML  oi  SM  SOIL LAYER sw  58  ii  *  11  2  o  o  ON  VO  in vo ;s;;S; m  2 ~ *  m  ©  ^  *  N o\  CN CN  *  °  i:i*:i:i  °  O  *  cn  3  !*  § '•' i - BBS  i t <L>  Vi  S- s f Bf  J  *  sSs;  o  CD O-  J  u  s s  3760  (180)  s;^i;  -£ -| |iii|i <^ :::S» S: ^ 1  Si S W :  ^jij|| -  u  ::  CD D) £Z CO JZ o o c CO  13.1  <N  (4.0)  (4.5-7.0)  22.9 - 39.3  8.2  (2.5)  14.7 - 22.9  15  ;;i«.  II  s  CO  2 ©C £> - ;;ss;tm  (7.0 -12.0)  9.8  (3.0)  4.9 - 14.7  (1.5-4.5)  4.9  -  y. "x. ssS;  •a | ^ 1 1  vm  62.2-75.3  ^ 00 v i * : 00  a  u  ::*:X  (19.0-23.0)  00  II  ON  :  23.0  CO  ii ^ o  :  (7.0)  00 .  x3 : :  39.3 - 62.2  45.9  (14.0)  e  -  CN  CO  3760  §111 l |  0-45.9 &  o  ;;.jj:  <N  5 a jigs;  A  1 ^ s  iiii ^  o  O  Sl-  o  (12.0 - 19.0)  ^ 111 ^ O  •tf  g  -a -  ° co  | : * : |  i;i*i;i;i  O CO  16.4  1 (0 - 14.0)  5222 29.5 -29.5 - 0  3  O  (1.5)  II  (250) iviviSi^^  ft(m)  1  j;.*;;; ©  iiiSilllii o  o  O  (9.0)  E ksf (MPa)  *  11° 111 * 11111 i^x  fi o  Depth ft(m)  Thickness  o  (180)  jijjjjjij  0-4.9  s  Layer  ?;s*#  a) E  o  •9k  °  °  o  (0-1.5)  c psf(kPa)  o  :  ?V s. g.  59  CHAPTER 6 LIQUEFACTION POTENTIAL ANALYSES For this section, two separate computer programs, 1D-LIQ and S H A K E , were employed to assess the liquefaction potential of saturated sand layers. As discussed earlier, these two programs employ significantly different methodologies:  1D-LIQ utilizes an effective stress  approach, and S H A K E utilizes a total stress approach.  Therefore, by employing these two  programs to assess the liquefaction potential at a site, the results from the two programs could be compared. L I Q U E F A C T I O N P O T E N T I A L A N A L Y S E S - 1D-LIO As a result of scope of work constraints and limitations of the 1D-LIQ program, only the preliminary soil profiles for the Tsawwassen and Ladner sites were modelled (refer to chapter 5).  The 1D-LIQ program is limited because it only uses the modulus and damping curves for  wet sand, wet clay and dry sand. It does not include modulus and damping curves for other soil types.  Consequently, only soil profiles which included these soils types could be modelled.  This resulted in only the Tsawwassen (north and south sides) and Ladner sites being modelled. At the Tsawwassen north site, a comparison of the excess pore pressure ratio versus depth for the 1:100 and 1:475 year events (refer to appendix figure CI) were made. A summary of the zones which are predicted to liquefy are provided in the following tables (refer to tables 19 and 20).  60 T A B L E 19.  T A B L E 20.  Summary of Liquefiable Zones for the 1:100 Year Event - Preliminary Soil Profile. Tsawwassen North (refer to appendix figure C D zone 1  0 to 18 feet (0 to 5.49 meters)  zone 2  78 to 84 feet (23.78 to 25.61 meters)  Summary of Liquefiable Zones for the 1:475 Year Event - Preliminary Soil Profile. Tsawwassen North (refer to appendix figure C D zone 1  0 to 18 feet (0 to 5.49 meters)  zone 4  78 to 84 feet (23.78 to 25.61 meters)  At Tsawwassen south, comparison figures of excess pore pressure ratio with depth were also made for soil profiles (refer to appendix figure CI). A summary of the liquefiable zones illustrated on the figures are provided in the following tables (refer to tables 21 and 22). T A B L E 21.  T A B L E 22.  Summary of Liquefiable Zones for the 1:100 Year Event - Preliminary Soil Profile. Tsawwassen South (refer to appendixfigureC D zone 1  0 to 24 feet (0 to 7.32 meters)  zone 2  38 to 44 feet (11.58 to 13.41 meters)  zone 3  48 to 64 feet (14.63 to 19.51 meters)  Summary of Liquefiable Zones for the 1:475 Year Event - Preliminary Soil Profile. Tsawwassen South (refer to appendixfigureC D zone 1  0 to 24 feet (0 to 7.32 meters)  zone 2  34 to 64 feet (10.37 to 19.51 meters)  zone 3  98 to 104 feet (29.88 to 31.71 meters)  At Ladner Exchange, comparison figures of excess pore pressure ratio versus depth were made for a single soil profile subject to 1:100 and 1:475 earthquake events (refer to appendix figure C2).  A summary of the potentially liquefiable zones illustrated on the figures are  61 provided in the following tables (refer to tables 23 and 24). T A B L E 23.  T A B L E 24.  Summary of Liquefiable Zones for the 1:100 Year Event - Prelim.nary Soil Profile, Ladner Exchange (refer to appendix figure C2) zone 1  0 to 5 feet (0 to 1.52 meters)  zone 2  15 to 25 feet (4.57 to 7.62 meters)  zone 3  45 to 60 feet (13.72 to 18.29 meters)  zone 4  98 to 108 feet (29.88 to 32.93 meters)  Summary of Liquefiable Zones for the 1:475 Year Event - Preliminary Soil Profile. Ladner Exchange (refer to appendix figure C2) zone 1  0 to 5 feet (0 to 1.52 meters)  zone 2  15 to 25 feet (4.57 to 7.62 meters)  zone 3  45 to 60 feet (13.72 to 18.29 meters)  zone 4  98 to 108 feet (29.88 to 32.93 meters)  LIQUEFACTION POTENTIAL ANALYSES - SHAKE The SHAKE analyses are divided into two parts: Liquefaction Potential Preliminary Analyses and Liquefaction Potential Detailed Analyses.  The preliminary analyses section  presents the results of the liquefaction assessment for all sites based on the preliminary soil profiles, level ground and the three preliminary earthquake records, with the exception of the North Tsawwassen site where four records were employed. In the detailed analyses, the detailed soil profiles from the Tsawwsassen and Ladner sites were used and a modified detailed soil profiles for these two sites. The modified soil profiles incorporated the presence of the existing embankment soils. Consequently, these analyses evaluated the liquefaction potential for both the level ground and embankment scenarios. In addition, the detailed analyses utilized a suite of 16 earthquake records.  62 L I Q U E F A C T I O N P O T E N T I A L A N A L Y S E S - S H A K E - Preliminary For the preliminary analyses, plots comparing the available in-situ (N ) values with the 1  60  required ( N ^ values were created for all five sites and are illustrated in appendix figures D l and D2. The required (NJ^Q values represented the actual required (N^^ value and mean plus one standard deviation was used. A summary of the potentially liquefiable zones outlined on the figures are provided in the following tables (refer to tables 25 and 26).  TABLE 25. Summary of Liquefiable Zones for the 1:100 Year Event - Preliminary Soil Profile. Tsawwassen North (refer to appendixfigureDl)  TABLE 26.  zone 1  0 to 24 feet (0 to 7.32 meters)  zone 2  28 to 48 feet (8.54 to 14.63 meters)  zone 3  58 to 62 feet (17.68 to 18.90 meters)  zone 4  78 to 84 feet (23.78 to 25.61 meters)  S u m m a r y of Liquefiable Zones for the 1:475 Year Event - Preliminary Soil Profile. Tsawwassen North (refer to appendixfigureDl)  zone 1  0 to 24 feet (0 to 7.32 meters)  zone 2  28 to 48 feet (8.54 to 14.63 meters)  zone 3  54 to 62 feet (16.46 to 18.90 meters)  zone 4  74 to 118 feet (22.56 to 35.98 meters)  At Tsawwassen south, plots were also made for soil profiles (refer to appendix figure Dl).  A summary of the potentially liquefiable zones illustrated in the figures are provided in  the following tables (refer to tables 27 and 28).  63  TABLE 27. Summary of Liquefiable Zones for the 1:100 Year Event - Preliminary Soil Profile. Tsawwassen South (refer to appendix figure Dl) zone 1  0 to 24 feet (0 to 7.32 meters)  zone 2  38 to 44 feet (11.58 to 13.41 meters)  zone 3  48 to 64 feet (14.63 to 19.51 meters)  zone 4  98 to 104 feet (29.88 to 31.71 meters)  TABLE 28. Summary of Liquefiable Zones for the 1:475 Year Event - Preliminary Soil Profile, Tsawwassen South (refer to appendix figure Dl) zone 1  0 to 24 feet (0 to 7.32 meters)  zone 2  28 to 34 feet (8.54 to 10.37 meters)  zone 3  38 to 64 feet (11.58 to 19.51 meters)  zone 4  74 to 104 feet (22.56 to 31.71 meters)  zone 5  108 to 124 feet (32.93 to 37.80 meters)  At the Ladner Exchange, plots were made for a single soil profile subject to 1:100 and 1:475 earthquake events (refer to appendix figure D2). A summary of the potentially liquefiable zones illustrated on the figures are provided in the following tables (refer to tables 29 and 30).  TABLE 29. Summary of Liquefiable Zones for the 1:100 Year Event - Preliminary Soil Profile. Ladner Exchange (refer to appendix figure D2) zone 1  0 to 5 feet (0 to 1.52 meters)  zone 2  15 to 25 feet (4.57 to 7.62 meters)  zone 3  60 to 64 feet (18.29 to 19.51 meters)  64  TABLE 30. Summary of Liquefiable Zones for the 1:475 Year Event - Preliminary Soil Profile, Ladner Exchange (refer to appendix figure D2) zone 1  0 to 5 feet (0 to 1.52 meters)  zone 2  15 to 35 feet (4.57 to 10.67 meters)  zone 3  45 to 64 feet (13.72 to 19.51 meters)  zone 4  88 to 98 feet (26.83 to 29.88 meters)  At Mathews Road Underpass, plots were also made for a single soil profile subject to 1:100 and 1:475 earthquake events (refer to appendix figure D3). A summary of the potentially liquefiable zones illustrated on the figures are provided in the following tables (refer to tables 31 and 32). However, it must be recognized that although the program predicts the development of cyclic stresses which would suggest liquefaction, it is unable to interpret the significance of the results.  Consequently when engineering judgement is applied, it would appear that the  triggering of liquefaction is unlikely in zone one because the soil is very organic and would not likely behave like a liquefiable granular soil.  TABLE 31. Summary of Liquefiable Zones for the 1:100 Year Event - Preliminary Soil Profile. Mathews Road Underpass (refer to appendix figure D3)  TABLE 32.  zone 1  10 to 18 feet (3.05 to 5.49 meters)  zone 2  38 to 42 feet (11.58 to 12.80 meters)  of Liquefiable Zones for the 1:475 Year Event - Prehminarv Soil Profile. Mathews Road Underpass (refer to appendixfigureD3) Summary  zone 1  10 to 18 feet (3.05 to 5.49 meters)  zone 2  38 to 42 feet (11.58 to 12.80 meters)  zone 3  52 to 76 feet (15.85 to 23.17 meters)  At Steveston Underpass, plots were also made for the single soil profile (refer to  65 appendix figure D4). Based on these figures, it appears that the proposed soil column does not contain any zones which would liquefy. At the Blundell Underpass, plots were made comparing the in-situ and required ( N ^ values for the 1:100 and 1:475 year earthquake events (refer to appendix figure D5). From this information, only one zone was identified which would liquefy during both the 1:100 and 1:475 year events (refer to appendix figure D5).  This zone was located at a depth of 24 to 28 feet  (7.32 to 8.54 meters). L I Q U E F A C T I O N POTENTIAL A N A L Y S E S - S H A K E - Detailed As discussed, only the Tsawwassen and Ladner sites were assessed, and a total of 36 earthquake records were utilized to evaluate the effects of different earthquake magnitudes, and A / V ratios.  In addition, at both of these sites, two different scenarios were studied.  One  scenario evaluated the existing level ground condition, while the second scenario assessed the condition in which an embankment has been placed ontop of the existing level ground. The level ground case was chosen because results from these analyses could be compared with the preliminary analyses, configuration.  and the general field conditions were best modelled using this  In addition, the embankment case was chosen because the effects on the  behaviour of the soil beneath the embankment could be evaluated. Unlike the detailed spectral response analyses, the liquefaction potential of a densified soil column was not assessed because it was assumed that a densified layer would not liquefy during the anticipated earthquake. This layer would not liquefy because the relative density of the soil layers would have been significantly increased and the drainage characteristics would have been greatly improved. The critical assumption in this reasoning, however, is that the densification technique used would  66 achieve these design assumptions. In addition, as in the preliminary analyses, only the liquefaction potential of the near surface soils were analyzed. However, in the detailed analyses, only the approximate top 92 feet (28.05 meters) and 79 feet (24.09 meters) were evaluated. In addition, the detailed analyses provides the required mean and mean plus one standard deviation (N ) values. Depending upon the degree of conservatism of the engineer, either the x  60  mean or mean plus one standard deviation required value could be used to compare with the insitu available ( N ^ value. Past practice has favoured the mean plus one standard deviation. However, current design practice (Byrne, P., 1994a) uses only the mean plus one standard deviation to evaluate the peak ground acceleration. If the mean plus one standard deviation was .  Using the only the average is preferred because it does not allow the development of  overconservatism into the design. Factors of safety are applied to determine the peak ground acceleration, and therefore, it is reasoned that added further factors of safety may be too conservative. For the 1:100 year and level ground scenario, the responses for the Tsawwassen site are shown in appendix figures E l and E2. In cases A (A/V = 1±0.2) and B (A/V = 1±0.5), soil layers between 6 and 39 feet (1.83 and 11.89 meters) were observed to liquefy.  The layer  between 39 and 62 feet (11.89 and 18.90 meters) was observed not to liquefy. However, the factor of safety was marginal. Beyond 62 feet or 18.90 meters, the factor of safety was approximately 1, which indicated that the soil could possibility liquefy. In case C, only one layer, located between 6 and 15 feet (1.83 and 4.57 meters) appeared to liquefy, based on the required ( N ^ value. The layer situated below this liquefiable layer, had a resistance which was  67 marginally greater than the required ( N ^ value. At depths greater than 23 feet (7.01 meters), the in-situ soil's resistance was greater than the required resistance. Consequently, layers below this depth were not considered to be liquefiable. In case D, the variation in the required ( N ^ values are significant, particularly between common recording stations.  Interestingly, the  variations are the largest between depths of 23 to 75 feet (7.01 to 22.57 meters). In general, all records indicated that the layers between 6 and 23 feet (1.83 and 7.01 meters) would liquefy. At greater depths, the liquefaction potential between records varied. However, the Griffith Park record (A/V = 0.88) generated the largest required (N^o values, while the Cal Tech record (A/V = 1.48) produced the smallest required ( N ^ values. Case E produced similar character for the required ( N ^ values as observed for cases A and B, except that the magnitude of required values was slightly reduced. Finally, in case F, two prominent zones were identified where it appeared that the soil would liquefy. The first zone was situated between depths of 6 and 23 feet (1.83 to 7.01 meters), while the second zone was located between depths of 39 and 92 feet (11.89 to 18.90 meters). Interestingly, this was the case where the liquefaction of the deeper layers was most probable. For the 1:475 event and Tsawwassen level ground condition, it was noticed that for cases A, B, D, and E, all of the soil layers analyzed liquefied (refer to appendix figures E3 and E4). In case C, only soil layers between 0 and 23 feet (0 and 7.01 meters) liquefied. Although not all of the soil layers liquefied, soil layers which did not liquefy were only marginally safe. Comparing these 1:475 event responses with case F (Mexico City, 0.05g) responses, it was observed that the magnitude of the required ( N ^ for the 1:475 responses was much larger, except when comparing cases C and F. In general, the largest difference between the required  68 values for cases A, B, D, and E, and case F occurred for the soil layers located between 0 and 39 feet (0 and 11.89 meters). For the 1:100 event and Ladner level ground scenario, it was observed that the soil layers assessed did not liquefy (refer to appendix figures E5 and E6). Interestingly, case B had the highest required (NJgo for the layer between 22 and 79 feet (6.71 and 24.09 meters), while case E had the highest required ( N ^ for the layers between 0 and 22 feet (0 and 6.71 meters). For the 1:475 event and Ladner level ground scenario, all cases generally indicated that the layer between 4 and 7 feet (1.22 and 2.13 meters) would liquefy (refer to appendix figures E7 and E8). In cases A, B, D and E, it was also recognized that between depths of 7 to 79 feet (2.13 to 24.09 meters) the required ( N ^ values were equal to or slightly greater than the in-situ (NJa, values. The other conditions analyzed considered the effects that the embankment would have on the liquefaction potential of underlying soils.  For the 1:100 event and the Tsawwassen  embankment case, the soil behaviour for the various cases studied are shown in appendix figures E9 and E10. It was noticed that in cases A through E, the average required (N^^ values did not exceed the in-situ values at any location. However, two layers were identified as being marginally safe. These two layers were situated between depths of 6 to 15 feet (1.83 to 4.57 meters) and 75 to 92 feet (22.87 to 28.05 meters).  In case F, the potential for liquefaction  between these same two layers, as referenced just previously, was much greater than for the preceding cases. In general all of the required ( N ^ values were significantly less than the values required for the level ground scenario. For the 1:475 event and the Tsawwassen embankment scenario, the general character of  69 the responses was similar to the responses for the 1:100 event except that the required ( N ^ values were slightly larger (refer to appendix figures E l l and El2). In cases A and B, two zones were identified where it was probable for liquefaction to occur. The first zone was located between depths of 6 and 23 feet (1.83 to 7.01 meters) while the second was situated between depths of 39 and 92 feet (11.89 to 28.05 meters). In case C, it was observed that no mean required ( N ^ value exceeded the in-situ values. However, the factor of safety against liquefaction was just slightly greater than one for the layer situated between depths of 6 and 15 feet (1.83 and 4.57 meters).  In case D, it was once again noticed that the Griffith Park  responses were generally larger than the Cal Tech responses. Nevertheless, in general, layers located between depths of 6 and 23 (1.83 to 7.01 meters) and 62 and 92 feet (18.90 to 28.05 meters) would likely liquefy. In case E, a slight reduction in the magnitude of the required (Ni) values as compared with cases A and B was observed. 60  In addition, the location of  liquefiable zones for cases A and B were similar to the zones recognized for case E. The only difference between the different cases was that for case E, the potential of liquefaction within these zones was slightly reduced.  Comparing the Mexico City response with the 1:475  responses, it was observed that the 1:475 event would be more likely to cause more soil layers to liquefy. For the 1:100 event and the Ladner embankment scenario, it was observed that in none of the cases did the soil layers appear to liquefy (refer to appendix figures E13 and El4). In addition, the smallest factor of safety against liquefaction was approximately 2.7, and this was observed for cases A and B. In general, the required ( N ^ values increased slightly with increasing depth.  70 For the 1:475 event and the Ladner embankment scenario, as in the previous scenario, the soil layers did not liquefy in any of the cases studied (refer to appendix figures E15 and E16). However, the smallest factor of safety against liquefaction was reduced to approximately 1.5.  Again, cases A and B, were the only two cases where the factor of safety against  liquefaction was this low.  In addition, the required ( N ^ values increased slightly with  increasing depth. The amplification of the required values also varied between cases. However, cases A and B appeared to exhibit the largest increase in the required values when comparing the 1:100 and 1:475 events.  71  CHAPTER 7 FLOW SLIDE STABILITY ANALYSIS - XSTABL Once liquefaction is triggered, the shear strength of the liquefied layer is significantly reduced. As a result, flow slides may develop, jeopardizing the stability of embankments. The potential of flow slide development was assessed by performing static analyses using the computer program X S T A B L , using an irregular surface Janbu stability analysis.  Based on  preliminary analyses information, the assessment consisted of inputting the relevant surface and sub-surface geometries, soil information, and ground water conditions. Flow slide assessments were carried out for only the 1:100 year event because of thesis scope limitations. The  embankment cross-section chosen for these analyses was a section situated  perpendicular to the highway alignment; the stability of the embankments in a section orientated parallel to the highway alignment was not assessed. Results from the preliminary liquefaction potential assessment were utilized to identify layers predicted to liquefy (refer to Chapter six).  Once predicted liquefiable layers were  identified, it was assumed that the shear strength within the liquefied layers could be defined by an estimated residual strength. Details regarding the determination of the residual strengths are provided in Chapter four and tables summarizing the soil parameters used in the analyses are provided in Chapter five. In addition, a summary of the residual strengths used can be observed in table 33. For these analyses, it was assumed that the surface crust for each site would not liquefy.  72  Site  Tsawwassen North Tsawwassen South Ladner Blundell Mathews Road  Zone (Location) of Liquefied Layer (depth below the Grd. Surface) ft(m) 14-18 (4.27 - 5.49) . 8 - 2 4 (2.43 - 7.32) 15-25 (4.57 - 7.62) 24-28 (7.32 - 8.54) 38-42 (11.59 - 12.80)  Thickness  (N,)60  ft(rn) 4 (1.22) 16 (4.88) 10 (3.05) 4 (1.22) 4 (1-22)  Residual Strength psf(kPa)  7 11 16 50 12  100 (5) 200 (10) 700 (35) 1500 (75) 300 (15)  TABLE 33. Summary of Residual Strengths Employed in the Flow Slide Assessment at the Sites Where Liquefaction Was Triggered (Tsawwassen, Ladner, Blundell. and Mathews Road) The following information summarizes the results of these analyses. For both the Tsawwassen North and South sites (refer to figures 9 and 10), flow slides are predicted when the near surface layers liquefy.  These results indicate that ground  improvement is necessary at these sites, if the embankments and bridge structure are to survive after liquefaction is triggered. Even if residual strength were optimistically increased by 5 kPa (100 psf) for both sites, the factor of safety for the north and south embankments was 1.063 and 0.678 respectively.  Indicating that flow slides would still likely be initiated in the area if  liquefaction were triggered. Therefore, remediation of the liquefiable soils at both sides of this embankment was necessary. As expected, the thickness of the liquefied zone had a significant effect in the initiation of flow slides.  This observation is illustrated when the stabilities of the North and South  Tsawwassen sites are compared (refer to figures 9 and 10). Stability results for the Ladner site, indicated that a flow slide would not develop as a  o CN  in  m m N"  CD CN CN CJ)  CO c CD  O  CD  to  C/3 C/3  co I/O  m CN to  o  LL.  O Q  TJ O  r>  CD CD  CQ <  o to C/J CN  X <  O  o  c  5  w  S  m  o f  i  z  .2 E CO c <  E  I CD  o  to  £s (7)  X  x:  CJ)  00 ^=  D  co  c  o CL T3 I  0  co W  CO  D 00  CO CD  0)  DC o  £ o  in CD  CO  CD i—  o CD CN  m CJ)  o  to  O^) SIXV-A  m IO  0_  £ O CO CO  CD  m CN m  2  O  75  76 result of sub-surface layers liquefying (refer to figure 11). However, as was discussed eariler, eventhough a flow slide is not predicted, significant deformations could still develop as a result of lateral spreading and settlement. Similar to the Ladner site, analyses at the Blundell and Mathews Road sites indicated that the embankments had factors of safety greater than one (refer to figures 12 and 13).  At the  Blundell site the most critical failure surface passes through the liquefied zone, but it appears that sufficient stronger material overlays this zone. At the Mathews Road site, the most critical failure surface does not even pass through the liquefied zone. Again, the liquefied zone at this site is located deep in the soil profile, and is overlain by stronger soils. Consequently, if the liquefied zone is located deep in the soil profile and strong soils overlay the liquefied, the probability of flow slides developing is minor. However, eventhough these results indicate that flow slides are not generated, it should be recognized that lateral spreading and settlement could still be a problem. In the subsequent section, deformations related to non-flow slide events are discussed.  77  78  (+990 S I X V - A  79  CHAPTER 8 DEFORMATIONS - DETAILED ANALYSES As discussed in Chapter 2, flow slides are not the only mechanism by which the ground distort. Significant deformations can develop by lateral spreading and settlement. The following provides a discussion of predicted horizontal ground deformations at the Tsawwassen site using three methods.  The methods used were the Newmark, LIQDISP (Extended Newmark), and  Finite Element Extended Newmark.  The settlement deformations were predicted using  Tokimatsu and 1D-LIQ methods. These methods are discussed in Chapter four. In addition, the reader is directed to Chapter five for a comprehensive discussion regarding the soil parameters used in this section.  For succinctness, the soil parameters used are not included  again in this chapter. DEFORMATIONS - Horizontal The Tsawwassen site was chosen and the 1:100 year event was assessed because at this site, it was predicted that soil layers underlying the embankment would liquefy during the anticipated 1:100 year earthquake event. In addition, the Tsawwassen site was determined to be the most critical site because several critical systems, such as the railway, gas line and hydroelectric power lines, were located close to the site and would likely be adversely effected by the embankment if it were to deform. The 1:100 year event was chosen because it was determined that similar subsurface layers would liquefy during both the 1:100 and 1:475 events. It was recognized that the velocities and accelerations would be larger for the 1:475 event than they would be for the 1:100 event. However, it was decided that the costs for remediation based on the 1:475 event would exceed available budgets whereas the remediation costs for a 1:100 event  80 would be within acceptable limits. In conducting these deformation analyses, it was decided to evaluate the effects of spot densification rather than densifying the whole liquefiable zone to limit horizontal displacements. By densifying zones surrounding and underneath the embankment, it was speculated that horizontal deformations could be reduced. These densified zones would act like dense soil plugs where liquefication would not occur. Therefore, failure planes which would develop through the liquefied zones and then through the densified zones, would be subject to significant resisting forces from within the densified zone.  When the resisting forces from the densification were  sufficient, horizontal deformations would be significantly reduced. In addition, it was hoped that the presence of the densified material beneath the embankment would support some of the mass of the embankment, and would thus reduce the amount of mass which would be contributing to the momentum driving forces. It was also speculated that if these densified zones were constructed employing such soil improvement techniques as stone columns or sand piles, the development of high pore pressures during the earthquake could be reduced. Research being conducted at the University of British Columbia (Byrne, P., 1992), has indicated that stone columns may act as conduits for dissipating pore pressures developed during an earthquake. Consequently, the stability of the embankment would be potentially increased as a result of the reduced liquefaction potential of adjacent soils and the more rapid return of the soil's stremgth as the excess pore pressure dissipates. As a result, soil densification, utilizing stone columns or sand piles, would contribute in several ways to the reduction of liquefaction and deformations. It should be recognized that the following analyses do incorporate the contributions that soil densification makes regarding  81 increased shear resistance and reduced momentum forces but do not include the effects of pore pressure dissipation. Utilizing the Tsawwassen site, an anticipated 1:100 year event earthquake condition, and five densification scenarios, the deformations were calculated by the Newmark, Byrne and finite element program (SOILSTRESS-DEF13) methods.  For these analyses, the depth of the  densification was held constant, while the width of the densified soil zone and the location of the densified soil zone, relative to the embankment, was varied. The depth of densification was held constant at 14.7 metres for the following reasons: 1) based on the previous analyses, it was determined that the soil layers located just below the crust would be the most likely to liquefy during the anticipated earthquake and therefore, the densification would not have to extend very far beneath the surface, and 2) based on experience obtained from soil densification work conducted at the 34B underpass site, Tsawwassen, and a review of available literature (Mitchell, J., and Wentz, F., 1991, Wrightman, A . , 1991, Barksdale, R., and Bachus, R., 1983, and Baumann, V . , and Bauer, G., 1974), it was decided that the practical limit of densification using the vibro-replacement is approximately 30 metres. 3) based on the liquefaction prediction information, the zone of liquefied material extended to a maximum depth of 14.7m, therefore densification below this depth was unnecessary. In the first scenario, a densified zone 16 metres wide was studied. Of the 16 metres, 6 metres of densification were located beneath the outer flanks of the existing embankment and  82 10 metres were situated around the perimeter of the embankment (refer to figure 14).  This  densification scheme was based upon the design used at the 34B Avenue underpass. The 34B Avenue underpass is located just north of the Tsawwassen site. Results from the three methods varied significantly (refer to table 34). The Newmark method predicted the smallest deformation while the LIQDISP method, for both a linear and non-linear modulus, predicted the the greatest displacement.  The displacement predicted by the F E M method was less than the LIQDISP  method, but the F E M calculated displacement was based on non-convergence of the energy term.  TABLE 34. Summary of Predicted Horizontal Deformations Based on Simplied and Detailed Analysis Methods Test  Newmark Method  (m)  Byrne Method (LIQDISP)  F E M Method  Linear Modulus Non-Linear Modulus (SOILSTRESS) (m) (m) (m) 10.5 24.1 18.1  I  0.146  II  0.032  8.2  10.9  8.5  III  0.032  7.8  10.3  7.5  IV  0.021  6.4  8.5  V  0.015  5.1  6.8  2.7 0.3  The Newmark method predicted displacements which were < 1 % of the LIQDISP results and approximately 1 % of the F E M result. Comparing the displacement predicted by the linear modulus method of LIQDISP and the non-linear modulus method of LIQDISP, it was interesting to note that the non-linear modulus displacement was greater and the ratio between the two (linear modulus/non-linear modulus) was 0.75.  In the F E M analysis, the energy term did not  converge within tolerable limits. Viewing the deformation plot generated by the F E M program (refer to Appendix F), it was noted that predicted embankment shape was substantially distorted.  Figure 14 Geometric Configuration of the Soil Layers and Embankment (Densified Soil Zone 41-57 m & 93-109 m : Test 1)  84 The distortion of the deformation plot, combined with the failure to achieve convergence in the energy term, suggests the development of significant deformations. For the second scenario, the width of the densified zone was increased to 20 metres. The width of the densified zone beneath the embankment flanks was increased from 6 to 10 metres, while the width of the densified zone around the perimeter remained at 6 metres (refer to figure 15). Once again, the Newmark method predicted the smallest displacement, while the non-linear modulus LIQDISP method predicted the greatest. Interestingly, the Newmark generated result was approximately 1 % of the other methods. Deformations predicted by the LIQDISP and FEM methods were very close. In particular the displacement predicted by the non-linear modulus LIQDISP method was very close to the FEM predicted displacement. Again, it was observed that the ratio between predicted displacements based on the linear modulus LIQDISP and nonlinear LIQDISP was 0.75. Comparing this densification scheme with the previous scheme, it was noted that increasing the densified zone beneath the embankment appeared to have a significant effect on reducing the displacements. In the third scenario, the width of the densified zone was increased to 23.5 metres. The portion of the densified zone around the perimeter was increased to 13.5 metres while the width of densified soil beneath the embankment remained at 10 metres (refer to figure 16). The displacement predicted by the Newmark method was once again very small, approximately 1 % of the displacement predicted by the other methods (refer to table 34).  The linear modulus  LIQDISP and FEM both predicted similar displacements, but were both slightly less than the displacement predicted by the non-linear modulus method. The ratio between the predicted  85  Figure 15 Geometric Configuration of the Soil Layers and Embankment (Densified Soil Zone 41-61 m & 89-109 m : Test 2)  86  Figure 16 Geometric Configuration of the Soil Layers and Embankment (Densified Soil Zone 37.5-61 m & 89-112.5 m : Test 3)  87 displacements based on the linear and non-linear LIQDISP methods was 0.75. Comparing this densification scheme with the previous scheme, it was noticed that only a marginal decrease in the predicted displacements was achieved when the width of densification around the perimeter was increased. In the four scenario, the area densified was completely confined to an area located underneath the flanks of the embankment footprint (refer to figure 17). The width of the area densified for this scenario was 14 metres.  It was decided that only the margin would be  densified because this would likely be the most accessible area. For this densification scheme, a significant variation between the three methods was observed (refer to table 34). Again, the Newmark method showed very small displacements which were in the order of only a few millimetres.  A displacement of only a few millimetres was not considered to be within  acceptable limits of certainty for displacement prediction and was consequently considered to be insignificant.  The FEM method predicted significant displacement but this predicted  displacement was substantially less than the displacements predicted by the LIQDISP methods. The linear modulus LIQDISP method was approximately a factor of 2.4 greater than the FEM displacement while the non-linear modulus LIQDISP method was approximately a factor of 3.1 greater. Once again, the ratio of the LIQDISP linear modulus and non-linear modulus predicted displacements was 0.75. All three methods predicted a significant reduction in the displacements, comparing this fourth scheme with the other schemes.  For example, the reduction between the fourth  densification scenario and the first scenario was approximately 65 % for the LIQDISP methods and 25% for the FEM method. It should be recognized that the zone of densification was 16  88  Figure 17 Geometric Configuration of the Soil Layers and Embankment (Densified Soil Zone 51-65 m & 85-99 m : Test 4)  89 metres for the first scenario and only 14 metres for the fourth scenario. The fifth densification scenario was similar to the fourth because the densification zone was located only beneath the footprint of the embankment. However, in the fifth densification scenario, the width of the zone densified was increased to 18 metres (refer to figure 18). Again the Newmark method predicted very minor displacements. In addition, both LIQDISP methods predicted significant displacements (refer to table 34).  However, the displacements were  reduced in magnitude from the previous scenario by approximately 20%. The ratio between the two LIQDISP methods also remained at 0.75. Comparing the FEM predicted displacement from the previous scenario with this scenario, it was noticed that the displacement had been reduced by approximately 90%. As stated previously, for brevity, the deformation patterns are provided in appendix G. In general, it was found that displacements were significantly reduced when the densified zone was located beneath the embankment. The greater the proportion of the embankment which was underlain by densified soil the better. The optimum densification scheme appeared to be densification scheme 5 which had an 18 metre wide densification zone on both flanks of the embankment. Of the three methods, the FEM method appeared to predict the most probable results. It was speculated that the FEM was the most successful at producing probable results because it was able to model the 2-dimensional problem. The Newmark method appeared to produce very small displacements, while the LIQDISP method appeared to only produce reasonable results for the first three densification schemes. The displacement predictions using the two LIQDISP methods were fairly close, with the non-linear method predicting displacements which were approximately 33 % greater then the linear method.  Figure 18 Geometric Configuration of the Soil Layers and Embankment (Densified Soil Zone 51-69 m & 81-99 m : Test 5)  91 DEFORMATIONS - Vertical Three methods were employed to determine the amount of settlement of the ground surface. One method was based on work carried out by Tokimatsu and Seed (Tokimatsu, K . , and Seed, H . , 1987). The second method was based on output provided by the computer program 1D-LIQ, and the third method was based on F E M results. The settlement estimations using the Tokimatsu and Seed, and 1D-LIQ methods were carried out for the five sites which had the preliminary soil profiles.  The F E M results were  obtained only for the detailed Tswwassen soil profile. The following section provides a summary of the estimated ground settlement at the five sites.  As discussed previously, the 1D-LIQ program was only used for analysing the soil  profiles at the Tsawwassen and Ladner sites. As a result, 1D-LIQ generated values are only available for those two sites. The 1D-LIQ calculated settlement predictions were, however, significantly influenced by liquefaction of horizons located well below the ground surface. It was noticed that some layers would settle to almost 50% of their thickness.  This resulted in estimated settlements which  ranged from 7 to almost 14 feet. The liquefaction of the deep seated layers was also dependant upon the duration of the earthquake. Longer duration earthquakes were generally responsible for causing the deeper layers to liquefy. Although the earthquake records were for the same event, the durations were different.  The run times of the 1D-LIQ program also  varied.  Consequently, the estimated settlements varied between each output record, and the variation was significant.  Due to the variation in estimated settlements, it was decided not to include  these settlements in the discussion. However, it should be noted that an error in the computer  92 code was identified and has since been corrected (Byrne, P., 1991). Table 35 provides a summary of the predicted settlements based on the Tokimatsu and Seed (Tokimatsu, K., and Seed, H . , 1987) methodology. These are estimates of the potential settlement and thus may be used as a guide. Interestingly, the ground settlements indicated by the F E M program were difficult to identify because the soil mass was deforming due to several factors.  However, once  densification beneath the embankment began to occur, and the horizontal displacements were significantly reduced, the identification of possible settlement induced deformation became apparent. In the fifth densification scheme (refer to figure 18), settlement was estimated to be responsible for approximately 0.1 metre of vertical displacement, and this displacement occurred primarily in the vicinity of the embankment core, the area where no densification had occurred. It must be stressed that this estimate is only an interpreted value and therefore is subject to discussion.  However, for the purposes of assessing the effectiveness of the densification, it  could be suggested that the densification beneath the embankment reduces the possiblity of the embankment settling after the earthquake.  TABLE 35. Estimated Average Settlements Based on the Tokimatsu and Seed. 1987. Procedure (These values are based solely upon the Tokimatsu and Seed Procedure (Tokimatsu and Seed, 1987). It should be noted that estimates from the 1D-LIQ computer program were as great as 7 ft for the Tsawwassen site and 17 ft for the Ladner site) SITE LOCATION  ESTIMATED SETTLEMENT  ESTIMATED SETTLEMENT  Tsawwassen North  1 ft (0.3 m)  1.5 ft (0.46 m)  Tsawwassen South  1 ft (0.3 m)  1.7 ft (0.52 m)  Steveston  < 0.1 ft (< 0.03 m)  < 0.1 ft (< 0.03 m)  Blundell  < 0.1 ft (< 0.03 m)  < 0.5 ft (< 0.15 m)  Ladner Exchange  1 ft (0.3 m)  1.5 ft (0.46 m)  Mathews Road  1 ft (0.3 m)  1.7 ft (0.52 m)  (1 in 100 year)  (1 in 475 year)  94  CHAPTER 9 REMEDIATION MEASURES Numerous densification techniques are available and could be employed to improve the ground stability at the designated sites. Restricted access at the sites, however, dramactically limits the equipment which could be used effectively.  Soil densification around the  embankments is also limited by the presence of services (e.x. waterlines, gaslines, hydrolines, etc.), roads or highways (unacceptable disruption of commuter traffic), existing bridge structure and private property. Densification around pile groups, which support the bridge, is hampered by limited head room as a result of the bridge deck. Consequently, ground improvement around the existing structures would be difficult and costly. Based on available information, several different soil improvement methods may be employed in order to densify liquefiable soils. The following is a list of some of the different methods which could be considered (Mitchell, J., 1993a):  A)  Dynamic Stabilization Techniques  - This method involves the compaction of cohesionless soils by inserting a vibrating cylindrical probe into the soils which causes the soil to vibrate and move into a closer particulate configuration.  For some of the methods, granular  material is backfilled around the probe as it is withdrawn. The placement of granular material effectively densifies the soil by displacement.  Occasionally,  water jetting at the tip of the probe is utilized in order to advance the probe. Examples of this technique are: vibrating probes (Terra or Phoenix probes),  vibro-compaction or vibro flotation, vibro-replacement, dynamic compaction and blasting. In addition to compacting the soil, this method creates granular columns within the soil mass. These granular columns typically have higher permeabilities than the surrounding soil mass and thus may act like pore pressure dissipation conduits. Therefore, this method has two advantages. B)  Compaction Piles  - This method involves the compaction of cohesionless soils by inserting nonstructural displacement piles (timber or pipe) or sand/gravel compaction piles into the desired soil. These piles effectively densify the soil by displacing the existing soil mass. C)  Grouting  - This method involves the densification of the soil mass by injecting materials into the soil voids. cement-soil-water  The injection material usually consists of a low slump  mixture.  Examples of this method include compaction  grouting, chemical grouting and jet grouting. D)  Mix-in-Place Piles and Walls  - This method consists of augering into the existing soil mass and mixing cement or chemicals into the soil in order to stabilize the soil mass.  It should be recognized that choosing a method of soil improvement, however, is influenced by several factors such as soil type, required depth of treatment, degree of soil densification required, permeability requirements, environmental concerns, availability of  96 equipment, headroom restriction construction access and project costs. Based on the Tsawwassen analyses, it was determined that effective remediation would require densification of the soils located beneath the embankment. By densifying beneath the embankment, the potential of lateral spreading and settlement, both due to liquefaction would be reduced. Densification of the soils beneath the embankment would be very difficult because the side slopes would make access for construction equipment difficult, the additional thickness of the soil due to the embankment side slopes would reduce the total thickness of soil that could densified, construction within the embankment side slopes would have to be carried out with due care in order to limit the potential of compromising the slope's stability, the densification process would have to be carried out to minimize the amount of settlement that it could generate, and the construction work would have to be contained in an area which would not disrupt commuter traffic.  As a result, it was decided that the most effective remediation method would be  compaction or chemical/cement grouting. However, this method has the two possible disadvantages of ground heaving and cost. It is expected that heaving of the ground could be controlled to within tolerable limits if the contractor is subject to significant cost penalties if heave was not controlled within designated levels. Based on conversations with Mr. B. Ninnes of Hay ward Baker (Ninnes, B., 1991), the cost of this technique would range from $12 to $15 per m . These costs were estimated during 3  the summer of 1991 and do not incorporate construction costs associated with working on the embankment slopes. Consequently, it was decided that an estimate of $15 per m would more 3  likely account for difficult working conditions and inflation. It was estimated that the total volume of material requiring densification (including the  97 area around the piles would be approximately 200,000 m for the Tsawwassen site. This volume 3  provides a rough estimate and would involve detailed volume calculations would be necessary once acceptable deformation criteria were established.  The deformation criteria would be a  function of determined acceptable deformation for the designated earthquake accerlation and velocity values. Based on the estimated volume, the cost of remediation would be approximately $3,000,000 to $5,000,000.  98  CHAPTER 10 CONCLUSIONS AND FUTURE RESEARCH This thesis presents the results of a geotechnical seismic assessment for five bridge sites located in Richmond and Delta.  The methodology employed provides a comprehensive and  effective means to assess a site. Based on this work, the following conclusions and discussion of future research are presented. CONCLUSIONS 1)  The accuracy of any analysis is dependant upon the input soil parameters and stress-strain relations. laboratory methods.  Soil parameters can be determined using in-situ and However, due to the difficulty and cost involved in  obtaining undisturbed field samples, determination of soil parameters in-situ is the most viable. Of the in-situ testing methods, interpreted SCPTU results provide the most accurate soil parameters, and should therefore be used where sands and fine grained soils exist. Laboratory results however, provide critical stress-strain relations which must be employed in order to predict confidently the soil behaviour.  In addition, laboratory testing is critical in providing the post  liquefaction residual strength. Consequently, both in-situ and laboratory testing is necessary to perform any seismic assessment. 2)  In order to assess the liquefaction potential of a soil column, S H A K E (one dimensional total stress method) and 1D-LIQ (one dimensional effective stress method) computer programs were employed. These programs showed agreement in their predictions of zones where liquefaction would develop.  However, of  these two, S H A K E program was favoured because it contained numerous soil modulus and damping reduction curves which are critical in modelling accurately the soil behaviour. For the five sites analyzed using S H A K E and ITER, it was determined that for the 1:100 and 1:475 events, soil layers would liquefy at the Tsawwassen Overpass, Ladner Exchange, Mathews Road Underpass, and Blundell Underpass, while no soil liquefaction would develop at the Steveston Underpass for either the 1:100 or 1:475 year event. However, it should be recognized that new K values M  have been presented (Loerscher, T . , and Youd, T . , 1994) which indicate that the Seed K values used in this analysis may be too conservative. M  Liquefaction of layers well below foundations may reduce the damage generated by the earthquake event. The liquefied zone may act as a base isolator, which may attenuate propagation of earthquake generated vibrations. A problem with relying upon these layers to reduce the damage at the surface is that they appear to liquefy only after the overlying layers have liquefied. Liquefaction is less likely to occur in densified zones where the density of the soil is increased or the areas beneath the embankment where the effective stress has been increased due the overlying embankment stresses. The presence of stone columns to densify the soil may also reduce the potential for liquefaction in adjacent soils because these stone columns could act as a conduit by which the excess pore pressures are permitted to dissipate. Prior to liquefaction, the soil deformations are generally small with shear strain  less than 1%.  However, once liquefaction is triggered, the soil stiffness and  strength are significantly reduced, the stress-strain behaviour is altered and the soil may undergo larege deformations due to the development of flow slides and/or lateral spreading and settlement. Therefore, significant deformations can develop and may be attributed to the development of flow slides, lateral spreading and settlement. The potential for flow slides was analyzed using a two dimensional limit equilibrium method (XSTABL) and the residual strength of the soil.  Based on  a review of available information, equation 25 was chosen to predict the residual strength of the liquefied zone, where S„ = oi a' s  vo  eqn. 33  the coefficient, a, was based on ( N ^ o values and limiting strains. Of the four sites that were identified where soil layers would liquefy for the 1:100 year event, it was only at the Tsawwassen site that a high probability existed for flow slides development. Even though flow slides may not develop, the potential exists that lateral spreading and settlement could develop.  Lateral spreading was modelled using  three methods which varied in their ease of use and degree of complexity. The three methods included the one dimensional Newmark, one dimensional extended Newmark (LIQDISP), and the two dimensional finite element with the extended Newmark energy theorem (SOILSTRESS - DEF13). The simple one dimensional Newmark was very limited because it used a rigid  plastic stress-strain model to predict soil behaviour and assumed the entire soil mass acted like a single block. Consequently, this method did not account for the significant deformations which could develop as the soil strains to reach the residual strength. This limitation was apparent when it was observed that the one dimensional extended Newmark and finite element methods predicted very much larger displacements.  In addition, being a one dimensional approach, the  displacement pattern could not be predicted, limiting the usefulness of this method limited. 11)  The one dimensional extended Newmark method (LIQDISP) predicted much larger displacements than the Newmark method, and only slightly larger than the finite element method for most cases. The largest variance between the LIQDISP and finite element methods was noticed for the best remediation scenario where the finite element method estimate was 10% of the LIQDISP method. However, LIQDISP has one significant limitation, in that it does not predict the deformation pattern.  Therefore, LIQDISP provides an excellent one dimensional cursory  method to rapidly predict displacements 12)  Liquefaction induced displacements were modelled using the finite element method with the extended Newmark energy theorem (SOILSTRESS - DEF13). This method was the most favoured because it was able to model accurately the post liquefaction soil stress-strain behaviour and incorporate inertia effects using the work energy balance theorem. In addition, this method allowed the prediction of displacement patterns and was easy to operate and use.  13)  Using these three methods, different remediation schemes were attempted at the Tsawwassen site. It was observed that the most effective remediation scheme was to densify a zone of soil directly beneath the embankment.  By instituting this  scheme, the finite element method predicted maximum displacements of less than 0.3 metres. 14)  The requirement that densification occur beneath the embankment limits remediation options.  Compaction or chemical grouting may be the only viable  remediation options. 15)  Compaction or chemical grouting, however, have two disadvantages: potential for ground heave and cost. Ground heave could be controlled by effective contract preparation and on-site monitoring. Cost would be high, but reflect the difficulty involved in dealing with the complex problems these sites pose.  16)  In addition, to horizontal displacements, recognized  as a potential  Tokimatsu/Seed method,  post liquefaction settlements are  design problem.  To predict settlements the  1D-LIQ method and SOILSTRESS-DEF13  finite  element method were employed. Settlement predictions by the Tokimatsu/Seed and finite element methods appeared reasonable, while the 1D-LIQ method predicted settlements which were as much as 50% of the layer thickness. Consequently, these 1D-LIQ settlement predictions were considered unlikely and should not be used. However, it has been recognized that the 1D-LIQ program has been corrected, and now predicted settlements are believed to be more accurate.  17)  Based on the Tokimatsu/Seed method, settlements at the Tsawwassen, Ladner and Mathews Road sites may be greater than 0.3 metres (approximately 1 foot). The Blundell and Steveston sites may also experience settlements, but the magnitude may be less significant.  Settlement across the site may also be significant as a  result of variations in soil stratigraphy and loading conditions. FUTURE RESEARCH 1)  Additional research into the influence of the type of earthquake (i.e. subduction versus strike-slip) on the response spectra and liquefaction potential.  2)  Additional research into comparing the results of the one dimensional total and effective stress programs with two dimensional total stress and effective stress programs regarding liquefaction potential.  3)  Additional research into determining the effects of using the Loertscher and Youd (Loertscher, T . , and Youd, T . , 1994) K  M  values to predict if liquefaction is  triggered at these five sites. 4)  Additional research into comparing the uncoupled deformation prediction methods with complete two dimensional effective stress methods.  5)  Additional field research into evaluating the effectiveness of stone columns to allow the dissipation of excess pore pressures during earthquake loading.  6)  Additional research into determining the problems associated with the one dimensional program 1D-LIQ in predicting settlements.  104 REFERENCES Alarcon, A . , and Leonards, G., 1988, "Discussion of Paper - Liquefaction Evaluation Procedure - by Poulos, Castro and France", Journal of Geotechnical Engineering, ASCE, Vol. 114, No. 2, p. 232-236 Anderson, D., Atukorala, U., Byrne, P., De Vail, R., Doyle, R., Gohl, B., Harrungton, E . , Lee, M . , Mill, A . , Nathan, N., Naesgaard, E . , Switler, B., Watts, B., and Weilier, G., 1991, "Earthquake Design in the Fraser Delta", Geotechnical Aspects Task Force Report, May 14, Arulanandan, K., and Muraleetharan, K., 1988, "Discussion of Paper - Liquefaction Evalution Procedure - by Poulos, Castro anf France", Journal of Geotechnical Engineering, ASCE, Vol. 114, No. 2, p. 236-239 Atukorala, U. 1991, Personal Conversation Baez, J., and Martin, G., 1993, "Advances in the Design of Vibro Systems for the Improvement of Liquefaction Resistance", Proceedings from the 7th Annual Vancouver Geotechnical Society Ground Improvement, Canadian Geotechnical Society, Vancouver, British Columbia Barksdale, R., and Bachus, R., 1983, "Design and Construction of Stone Columns: Volume I", United States Department of Transportation, Federal Highway Administration, FHWA/RD83/026, Washington D.C. 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M . , and Yan, L . , 1990, "1D-LIQ: A Computer Program for One-Dimensional Effective Stress Dynamic Analysis of Soil Layers", Soil Mechanics Series #146, Department of Civil Engineering, The University of British Columbia Byrne, P.M., Cheung, H . , and Yan, L . , 1987, "Soil Parameters for Deformation Analysis of Sands", Canadian Geotechnical Journal, Vol. 24, No. 3, p.366-376 Byrne P.M., Cheung, H . , and Yan, L . , 1986, "Soil Parameters for Deformation Analysis of Sands", Soil Mechanics Series #104, Department of Civil Engineering, The University of British Columbia Byrne, P . M . , and Atukorala, U . , 1985, "A Report to the Ministry of Transportation and Highways on Seismic Stability of Bridge Abutments Underlain by Organic Soils, Annacis Island Project", Department of Civil Engineering, University of British Columbia Byrne, P.M., Campanella, R., Vaid, Y., Atukorala, U . , and Gillespie, D., 1984, "A Report to the Ministry of Transportation and Highways on Insitu and Laboratory Testing, Annacis Island Bridge Project, Department of Civil Engineering, University of British Columbia Bynre, P.M., and Atukorala, U . , 1983, "Prediction of P-Y Curves From Pressuremeter Tests and Finite Element Analyses, Soil Mechanics Series #66, Department of Civil Engineering, University of British Columbia Byrne, P . M . , and Janzen, W., 1981, "SOILSTRESS: A Computer Program for Nonlinear Analysis of Stresses and Deformations in Soil", Soil Mechanics Series #52, Department of Civil Engineering, The University of British Columbia, updated 1989 Campanella, R., Hitchman, R., and Hodge, W., 1990, "New Equipment for Densification of Granular Soils at Depth", Canadian Geotechnical Journal, Vol. 27, p. 167-176 Campanella, R., Robertson, P., and Gillespie, D., 1986, "Seismic Cone Penetration Test", Proceedings of the In Situ '86 Specialty Conference, ASCE, Geotechnical Special Publication No. 6, p. 116-130 Clough, G., and Chameau, J., 1983, "Seismic Response of San Francisco Waterfront Fills", Journal of Geotechnical Engineering, ASCE, Vol. 109, No. 4, p. 491-506  107 CSMIP - California Strong Motion Instrumentation Program, 1989, California Department of Conservation, Division of Mines and Geology, Office of Strong Motion Studies, Sacramento, California De Alba, P., 1988, "Discussion of Paper - Liquefaction Evaluation Procedure - by Poulos, Castro, and France", Journal of Geotechnical Engineering, ASCE, Vol. 114, No. 2, p. 239-241 Dennis, N., 1986, "Discussion of Paper - Liquefaction Evaluation Procedure - by Poulos, Castro and France", Journal of Geotechnical Engineering, ASCE, Vol. 114, No. 2, p. 241-243 Dikmen, S., and Ghaboussi, M . , 1984, "Effective Stress Analysis of Seismic Responses and Liquefaction: Theory", Journal of Geotechnical Engineering, ASCE, Vol. 110, No. 5, p. 628658 Dumas, J., and Beaton, N., 1992, "Dynamic Compaction Suggested Guidelines for Evaluating Feasibility - for Specifying - for Controlling", Proceedings from the 45th Canadian Geotechnical Conference, Toronto, Ontario Duncan, J., Byrne, P., Wong, K., and Mabry, P., 1980, "Strength, Stress-Strain, and Bulk Modulus Parameters for Finite Element Analysis of Stresses and Movements in Soil Massess", Geotechnical Engineering Report No. UCB/GT/80-01, College of Engineering, University of California, Berkeley, California Ersoy, T., 1994, Personal Conversation Ersoy, T., 1993, Personal Conversation Ersoy, T., 1992, Personal Conversation Ersoy, T., 1991a, Personal Conversation Ersoy, T., 1991b, "Criteria for Geotechnical Assessment of Existing Infrastructure and New Design with Respect to Seismic Stability", Draft Copy of An Internal Ministry of Transportation and Highways of British Columbia Memorandum Esrig, M . , and Bachus, R., (editors), 1991, "Deep Foundation Improvements: Design, Construction, and Testing, STP 1089", American Society for Testing and Materials, Pennsylvania Finn, W.D.L., 1993, "Evaluation of Liquefaction Potential", Proceedings of the Seminar on Soil Dynamics and Geotechnical Earthquake Engineering, A.A. Balkema, p. 127-157 Finn, W.D.L., 1988, "Dynamic Analysis in Geotechnical Engineering", Proceedings of Earthquake Engineering And Soil Dynamics II - Recent Advances in Ground-Motion Evaluation  108 Specialty Conference, ASCE, Geotechnical Special Publication #20, p. 523-591 Finn, W.D.L., Woeller, D.J., and Robertson, P.K., 1989, "In Situ Determination of Liquefaction Potential and Dynamic Soil Properties: A Regional Study in Richmond, B.C.", Earthquake Geotechnical Engineering, Special Volume on Influence of Local Soils on Seismic Response, 12th International Conference on Soil Mechanics and Foundation Engineering, Rio de Janeiro, Brazil, p. 135-142 Finn, W.D.L., and Yogendrakumar, M . , 1988, "Probability of Seismically Liquefaction", Department of Civil Engineering, University of British Columbia  Induced  Finn, W.D.L., Yogendrakumar, M . , Yoshida, N., and Yoshida, H . , 1986, "TARA-3: A Program to Compute the Response of 2-D Embankments and Soil-Structure Interaction Systems to Seismic Loadings", Department of Civil Engineering, University of British Columbia Finn, W.D.L., Lee, K., and Martin, G., 1977, "An Effective Stress Model for Liquefaction", Journal of the Geotechnical Engineering, ASCE, Vol. 103, p. 517-533 Finn, W.D.L., Byrne, P., and Martin, G., 1976, "Seismic Response and Liquefaction of Sands", Journal of Geotechnical Engineering, Vol. 102, No. 8, p. 841-856 Golder Manual For ITER (PC Version), 1990, Golder Associates Hamada, M . , Towhata, I., Yasuda, S., and Isoyama, R., 1987, "Study of Permanent Ground Displacement Induced by Seismic Liquefaction", Computers and Geotechnics, #4, p. 197-220 Harder, L . F . , and Seed, H.B., 1986, "Determination of Penetration Resistance for CoarseGrained Soils Using the Becker Hammer Drill", UCB/EERC Report No. 86/06, University of California, Berkeley Harder, L . , Hammond, W., and Ross, P., 1984, "Vibroflotation Compaction at Thermalito Afterbay", Journal of Geotechnical Engineering, ASCE, Vol. 110, No. 1, p. 57-70 Hardin, B., and Drnevich, V . , 1972, "Shear Modulus and Damping in Soils: Desing Equations and Curves", Soil Mechanics and Foundations Division, Proceedings of the American Society of Civil Engineerings, ASCE, SM7, p. 667-692 Hunt, R., 1986, Geotechnical Engineering Techniques and Practices. McGraw-Hill, New York Idriss, I., 1990, "Response of Soft Soil Sites During Earthquakes", Proceedings of the 42 Canadian Geotechnical Conference, Winnipeg, Manitoba, p. 216-222 Idriss, I., 1985, "Evaluating Seismic Risk In Engineering Practice", International Conference SMFE IX, p. 255-320  109 Jinto, H . , 1993, Personal Conversation Jinto, H . , 1992, Personal Conversation Jinto, H . , and Byrne, P.M., 1994, "A Procedure for Predicting Seismic Deformation of Earth Structures", Settlements 94, Texas A & M, ASCE Specialty Conference Jong, H . , 1988, "A Critical Investigation of Post-Liquefaction Strength and Steady-State Flow Behaviour of Saturated Soils", Ph.D., Thesis, Department of Civil Engineering, Stanford University, California Kern, B., 1991, Personal Conversation Kutter, B., 1988, "Discussion of Paper - Liquefaction Evaluation Procedure - by Poulos, Castro, and France", Journal of Geotechnical Engineering, ASCE, Vol. 114, No. 2, p. 243-246 Loertscher, T., and Youd, T., 1994, "Magnitude Scaling Factors For Analysis of Liquefaction", Japan - USA Seismic Workshop Liao, S., and Whiteman, R., 1986, "Overburden Correction Factors for SPT in Sand", Journal of Geotechnical Engineering, ASCE, Vol. 112, No. 3, p. 373-377 Lister, D., 1991, Personal Conversation Luternauer, J., 1991a, Personal Conversation Luternauer, J., 1991b, "Geoarchitectural, Evoluation, and Seismic Risk Assessment of the Southern Fraser River Delta, B.C.", Preliminary Information (project status report, reflection seismic records, and deep drill hole data) from the Cordilleran and Pacific Geoscience Division, Vancouver, British Columbia MacFarlane, I., 1969, "Muskeg Engineering Handbook", Muskeg Subcommittee of the National Research Counsel, Associate Committee on Geotechnical Research, University of Toronto Press Mayne, P., and Kulhawy, F., 1982, "K -OCR Relationships in Soil", Journal of Geotechnical Engineering, ASCE, Vol. 108, No. 6, p. 851-872 D  McCammon, N., Fitzell, T., Atukorala, U. and Momenzadeh, M . , 1990, "Report to the B.C. Ministry of Transportation and Highways on Geotechnical Investigation and Response of Foundations to Seismic Loading at Vedder Canal Bridges (#533 and #1658) on Highway #1, British Columbia", Golder Associates Limited, Consulting Engineers, Vancouver, Contract T & H 1823 Mitchell, J., 1993a, "Ground Treatment for Seismic Stability of Bridge Foundations",  110 Proceedings from the 7th Annual Vancouver Geotechnical Society - Ground Improvement, Canadian Geotechnical Society, Vancouver, British Columbia Mitchell, J., 1993b, "Mitigation of Ground Failure Risk - Some Lessons From the Loma Prieta Earthquake, Ground Treatment for Seismic Stability of Bridge Foundations", Proceedings from the 7th Annual Vancouver Geotechnical Society - Ground Improvement, Canadian Geotechnical Society, Vancouver, British Columbia Mitchell, J., and Wentz, F., 1991, "Performance of Improved Ground During the Loma Prieta Earthquake", Earthquake Engineering Research Center Report, No. UCB/EERC 91-12, College of Engineering, University of California, Berkeley, California National Building Code of Canada, 1990 National Building Code of Canada Supplemental, 1991 Naesgarrd, E . , and Beaton, N., 1993, "Dynamic Compaction Densification for Liquefaction Mitigation and Improved Foundation Support in the Fraser Delta - A Case History", Proceedings from the 7th Annual Vancouver Geotechnical Society - Ground Improvement, Canadian Geotechnical Society, Vancouver, British Columbia Newmark, N., 1965, "Effects of Earthquakes on Dams and Embankments", Geotechnique, Vol. 15, No. 2, p. 139-160 Ninnes, R., 1991, personal conversation O'Rourke, T., Stewart, H . , Blackburn, F., and Dickerman, T., 1990, "Geotechnical and Lifeline Aspects of the October 17, 1989 Loma Prieta Earthquake in San Francisco", Technical Report NCEER 90-0001, Department of Structural Engineering, School of Civil and Environmental Engineering, Cornell University, New York Pacific Geoscience Centre, Acceleration and Velocity Information for Ladner and the George Massey Tunnel Pilecki, T., 1988, "Discussion of Paper - Liquefaction Evaluation Procedure - by Poulos, Castro and France", Journal of Geotechnical Engineering, ASCE, Vol. 114, No. 2, p. 246-247 Pillai, V . , and Stewart, R., 1993, "Evaluation of Liquefaction Potential of Foundation Soils at the Duncan Dam", Proceedings of the 46th Annual Canadian Geotechnical Conference, Saskatoon, Saskatchewan, p. 247-258 Poulos, S., Castro, G., and France, J., 1988, "Closure to the Discussion of the Paper Liquefaction Evaluation Procedure - by Poulos, Castro, and France", Journal of Geotechnical Engineering, ASCE, Vol. 114, No. 2, p. 215-259  Ill Poulos, S., Castro, G., and France, J., 1985, "Liquefaction Evaluation Procedure", Journal of Geotechnical Engineering, ASCE, Vol. I l l , No. 6, p. 772-792 Purssell, T., 1985, "Modulus Reduction Dynamic Analysis", Master Thesis, University of British Columbia, Department of Civil Engineering Pyke, R., 1988, "Discussion of Paper - Liquefaction Evaluation Procedure - by Poulos, Castro and France", Journal of Geotechnical Engineering, ASCE, Vol. 114, No. 2, p. 247-250 Quinn, G.A., and Stilley, A . N . , 1989, "Compaction Grouting to Improve Liquefiable Dam Foundation", Engineering Geology and Geotechnical Engineering, Editor Watters, Balkema, p. 189-196 Radloff, B, 1991, Personal Conversation Robertson, P., 1990, "Seismic Cone Penetration Testing for Evaluating Liquefaction Potential", Proceedings, Symposium on Recent Advances in Earthquake Design Using Laboratory and In Situ Tests, ConeTec Investigations Limited, Burnaby, British Columbia Robertson, P., 1982, "In-situ Testing of Soil With Emphasis on Its Application to Liquefaction Assessment", Ph.D., Thesis, Department of Civil Engineering, The University of British Columbia Robertson, P., and Campanella, R., 1989, "Guidelines for Geotechnical Design Using CPT and CPTU", Soil Mechanics Series No. 120, Department of Civil Engineering, The University of British Columbia Robertson, P., and Campanella, R., 1986a, "Guidelines for Use, Interpretation and Application of the CPT and CPTU", Soil Mechanics Series No. 105, Department of Civil Engineering, The University of British Columbia Robertson, P., Campanella, R., Guillespie, D., and Rice, A . , 1986b, "Seismic CPT to Measure In Situ Wave Velocity", Journal of Geotechnical Engineering, ASCE, Vol. 112, No. 8, p. 791803 Robertson, P., and Campanella, R., 1985, "Liquefaction Potential of Sands Using the CPT", Journal of Geotechnical Engineering, ASCE, Vol. I l l , GT 3, p. 384-403 Robertson, P., and Hughes, J., 1985, "Determination of Properties of Sand From Self-Boring Pressuremeter Tests", Soil Mechanics Series No. 90, Department of Civil Engineering, The University of British Columbia Robertson, P., Campanella, R., and Wightman, 1983, "SPT-CPT Correlations", Journal of Geotechnical Engineering, ASCE, Vol. 109, GT 11, p. 1449-1459  112 Rollins, K., and Seed, H . , 1990, Influence of Buildings on Potential Liquefaction Damage", Journal of Geotechnical Engineering, ASCE, Vol. 116, No. 2, p. 165-185 Salgado, F., and Pillai, V . , 1993, "Seismic Stability and Deformation Analysis of Duncan Dam", Proceedings of the 46th Annual Canadian Geotechnical Conference, Saskatoon, Saskatchewan, p. 259-270 Schnabel, P., Lysmer, J., and Seed, H., 1972, "SHAKE - A Computer Program for Earthquake Response Analysis of Horizontally Layered Sites", Report #EERC 72-12, College of Engineering, University of California, Berkeley, California Seed, H . , 1987, "Design Problems in Soil Liquefaction", Journal of Geotechnical Engineering, ASCE, Vol. 113, No. 8, p. 827-845 Seed, H . , and De Alba, P., 1986, "Use of SPT and CPT Test for Evaluating the Liquefaction Resistance of Soils", Proceedings of the In Situ '86 Specialty Conference, ASCE, Geotechnical Special Publication No. 6, p. 281-302 Seed, H . , Wong, R., Idriss, I., and Tokimatsu, K., 1986, "Moduli and Damping Factors for Dynamic Analyses of Cohesionless Soils", Journal of Geotechnical Engineering, ASCE, Vol. 112, No. 11, p. 1016-1033 Seed, H . , Tokimastu, K., Harder, L . , and Chung, R., 1984, "The Influence of SPT Procedures in Soil Liquefaction Resistance Evaluations", Earthquake Engineering Research Center Report, No. UCB/EERC 84-15, College of Engineering, University of California, Berkeley, California Seed, H . , Idriss, I., and Arango, I., 1983, "Evaluation of Liquefaction Potential Using Field Performance Data", Journal of Geotechnical Engineering, ASCE, Vol. 109, No. 3, p. 458-482 Seed, H . , and Idriss, I., 1982, "Ground Motions and Soil Liquefaction During Earthquakes", Earthquake Engineering Research Institute Seed, H . , 1979, "Soil Liquefaction and Cyclic Mobility Evaluation for Level Ground During Earthquakes", Journal of Geotechnical Engineering, ASCE, Vol. 105, No. 2, p. 210-255 Seed, H . , and Idriss, I., 1970, "Soil Moduli and Damping Factors for Dynamic Response Analyses", Earthquake Engineering Research Center Report, No. EERC 70-10, College of Engineering, University of California, Berkeley, California Seed, R.B., and Harder, L . F . , 1990, "SPT-Based Analysis of Cyclic Pore Pressure Generation and Undrained Residual Strength", Proceedings of the H. Bolton Seed Memorial Symposium, Bitech Publishers, p. 351-376 Sharma, S, 1990, "XSTABL:  An Integrated Slope Stability Analysis Program for Personal  113 Computers, Version 3.20", Interactive Software Designs Inc., Moscow, Idaho Skempton, A . , 1986, "Standard Penetration Test Procedures and the Effects in Sand of Overburden Pressure, Relative Density, Particle Size, Ageing and Overconsolidation", Geotechnique, Vol. 36, No. 3, p. 425-447 Srithar, 1991, Personal Conversation Stokoe, K., and Hoar, R., 1978, "Variables Affecting In Situ Seismic Measurements", Proceedings of the Earth Engineering and Soil Dynamics, Specialty Conference, ASCE, Vol. 2, p. 919-939 Surfical Geology - Vancouver, Map 1486A, Geological Survey of Canada, 1979 Surfical Geology - New Westminster, Map 1484A, Geological Survey of Canada, 1980 Sy., A . , Henderson, P., Lo, R., Siu, D., Finn, W., and Heidebrecht, A . , 1991, "Ground Motion Response for Fraser Delta, British Columbia", Fourth International Conference on Seismic Zonation, Stanford California Tao, S, 1991, Personal Conversation Telford, W.M., Geldart, L.P., Sheriff, R.E., and Keys, D.A., 1976, Applied Geophysics. Cambridge University Press, Cambridge, p. 218-261 To, P., and Broomhead, D., 1993, "Ground Stabilization by Compaction Grouting - A Case History", Proceedings from the 7th Annual Vancouver Geotechnical Society - Ground Improvement, Canadian Geotechnical Society, Vancouver, British Columbia Tokimatsu, K., and Seed, H . , 1987, "Evaluation of Settlements in Sands Due to Earthquake Shaking", Journal of Geotechnical Engineering, ASCE, Vol. 113, No. 8, p. 861-878 Watts, B., and Konrad, J., 1993, "Decision Analysis for Ground Improvement in Fraser River Delta" Proceedings from the 7th Annual Vancouver Geotechnical Society - Ground Improvement, Canadian Geotechnical Society, Vancouver, British Columbia Vaid, Y.P., Chung, E.K.F., and Kuerbis, R.H., 1989, "Stress Path and Steady State", Soil Mechanics Series No. 128, Dept. of Civil Engineering, University of British Columbia Vaid, Y.P., and Chern, J.C., 1985, "Cyclic and Monotonic Undrained Response of Saturated Sands", ASCE National Convention, Session - Advances in the Art of Testing Soils Under Cyclic Loading, p. 120-147 Wrightman, A . , 1991, "Ground Improvement by Vibrocompaction", Geotechnical News, Vol.  114 9, #2, p. 39-41 Y u , P. and Richart, F . , 1984, "Stress Ratio Effects on Shear Modulus of Dry Sands", Journal of Geotechnical Engineering, A S C E , Vol. 110, No. 3, p. 331-345  115  APPENDIX A  116  U  B  I N  C  Slta  S I "r u  Locatlora GILLESPIE F i l m 1 acinar. «crt C o n * U w c h SUPER  FRICTION RATIO Rf CD  T On  CONE BEARING Oe (MPa)  E  S  T  I  N  G  CPT D a t a i R 8 1 / 1 2 / 1 1 1 3 . 3 0 S i to Loot LADNER EXCH SE Coanrantsi BRANCH INTERPRETED PROFILE  PORE PRESSURE U O l of vatar)  0  80  a L  ai ai  E UJ a  20-  Oapfeh I n e r w n n t •  .OS  •  Mem O a p t h i  2 0 . 85  •  117  U B C SIto  I M  Location* Con*  S I "r u ' T E S T I N G CPT D a t a t  GILLESPIE  Film  l a d n a r . «jc±t  Uoadt  SUPER  On  R81/12/11  13.30  L o o . LADNER EXCH SE  Co«aant«« CONE BEARING Oc 0Pa>  FRICTION RATIO Rf CD  SIt«  BRANCH INTERPRETED PROFILE  PORE PRESSURE U Cm. of watur) 20  3CH  4t> Depth  Inerwamnt  •  .05  •  Hem D a p t h i  20. 85  •  118  APPENDIX B  119 T  S  A  Englnaar • GILLESPIE CPT i T6V001 P Cans U N d • SEISMIC  f l  W  A  1 / 3  3  E  M  CPT Oat* • 91/12/11 00*38 L o c a t i o n • T8V NORTH VEST E l a v a t i o n • BRANCH IMIUPHLILU PKOFIU  T V HESCTAICE  RUCTION RATIO Fa/a CD a . . . . 9  OB OTO)  1*  XS4  a> Oapth  I  .05 •  Mem Oapth i  48.43 *  120 T Engineer  t  CPT i Cone Ueed t FRICTION M T U P*VB OD  S  A  W  A  S  GILLESPIE TSV001  Pg 2 /  3  SEISMIC  S  E  N  CPT D a t e •  81/12/11  Location t  TSV NORTH VEST  Elevation t TIT RESISTANCE Oo Ot°u>  Depth Inereeerrt t  .OS •  08,33  BRANCH  PORE PRESSURE P» (bar)  Max D e p t h •  WT0RPRETEO PROFILE  40.43  •  3 Englnaar  •  CPT • t FRICTION RATIO Fa/8 CO  A  W  A  3  GILLESPIE TSV001  Pg 3 /  3  EI  N  CPT D a t a i  81/12/11  Location i  TSW NORTH WEST  Elavation •  SEISMIC  08*39  BRANCH DdWRtlEfl PROFILE  TIP RESISTANCE Oa O N  40  45H  O  9  L 9  J  304  I tQ.  UJ  a  a> Oapth I n o r — a n t  t  .OS  •  Max O a p t h •  48, 43  •  [  122  APPENDIX  C  123  C1 a Tsawwassen North (1:100 Year)  Depth (ft)  150  C1 b Tsawwassen South (1:100 Year)  75  150  Depth (ft)  C1 c Tsawwassen North (1:475 Year)  75  150  Depth (ft)  C1 d Tsawwassen South (1:475 Year)  Depth (ft)  150  Figure C1 IDENTIFICATION OF POTENTIALLY LIQUEFIABLE ZONES FOR 1:100 and 1:475 YEAR EARTHQUAKES AT THE TSAWWASSEN SITE (SOIL PROFILE BASED ON OLD SPT INFORMATION)  124  C2a Ladner (1 MOO Year)  75  150  Depth (ft)  C2b Ladner (1:475 Year)  75 Depth (ft)  Figure C2 IDENTIFICATION OF POTENTIALLY LIQUEFIABLE ZONES FOR 1:100 and 1:475 YEAR EARTHQUAKES AT THE LADNER SITE (SOIL PROFILE BASED ON OLD SPT INFORMATION)  150  125  A P P E N D I X  D  /"^k:  n 30  i  i Required (N1)60-Grlfflth Park Required (N1J60-Cal Tech Required (N1)60;Hughes Lake 12 Required (N1J60-Hughes 4 Available (N1)60  j.  I i  126  D1 a Tsawwassen North (1:100)  60  SF •-  1  ....V  ^t;--  /  ..../ /  /  A\ "/ < \  °  ; \  1 \ i \  *"  >  1  ;.'  t 20  eo 10  40  i  ,/"*\ 0--..  ::  j  \  ••-g-  H*,... \  ••V2;-.  p-vrj^;  60  \  V  ^  ]  •  !=--»—*  80  100  ; \  120  140  Depth (tt)  0  D1 b Tsawwassen South (1:100) Required (N1)60-Grlfffih Park Required (NV)60-Cal Tech Required (N1)60-Hughes Lake 12 Available (N1)60  140  D1 c Tsawwassen North (1:475)  60  i  •  Required (N1)60-Grlfflth Park Required (NrjeO-TJal Tech Required (NIlKKHughes Lake 12 Required (N1)60-Hughes Lake 4 Avallabie(N1)60  40  5  /  x  \  /  i  \  3 0  ,x*si y /J* O• * :\ \  • —  Z.20  o —  O \"<>  n -  t OT 10  ^—  ~~~~„.  1  \ ' •  » * >:.:.-:--.:.:.:0... j  20  40  eo  60  100  120  140  Depth (ft)  D1 d Tsawwassen South <1:475) Required (N1)60-Grlfflth Park Required (Nrj~60-Cal Tech Required (N1)60+iughes Lake 12 Available (N1)60  Depth (  Figure D1 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1 )60 VALUES FOR THE 1:100 AND 1:475 EVENTS (TSAWWASSEN SITE, NORTH AND SOUTH) (SOIL PROFILE BASED ON OLD SPT INFORMATION)  127  D2a Ladner (1:100 Year)  35  I  ^30  *  ^25  /i Y  j 20  \  8 15 &10 l-  a.  -i  I  Required (N1)60-Griffith Park Required (N1)60-Cal Tech Required (N1)6CMHughes Lake 12 Available (N 1)60  •/ j  .t^ba A  |  ^Y  W 5  20  40  \  ./..j ^fl,^. W | -A-60  80 Depth (ft)  \  "—— it 100  120  \  \  \  1 140  160  D2b Ladner (1:475 Year)  80 Depth (ft)  160  Figure D2 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR THE 1:100 AND 1:475 EVENTS (LADNER SITE) (SOIL PROFILE BASED ON OLD SPT INFORMATION)  128  D3a Mathews Road (1:100 Year) 120  f  I  I  Required (N1)60-Griffith Park 1M  Required (N1)60-Cal T e c h  % o .Q X  8  60  Z  40  §5  20  Required (N1)60-Hughes Lake 12  80  Available (N1)60  t  I .A.  20  /\ D oI a A™ . -.  ••• a  h  40  60  .*' ;:.r.r.:.:.--r.fi { I .  .A....  80  .--:.ft •  100  120  140  I  I  160  Depth (ft)  1  D3b Mathews Road (1:475 Year) 120  I  Required (N1)60-G riff ith Park 100  Required (N1)60-Cal T e c h  % o  80  ~o  60  n  Required (N1)60-Hughes Lake 12 Available (N1)60  CO  40 §5  >"/-Q\-&  20  V7"" " '\-& Q  — •A..  J  -^..-.^ .^...1 <  20  40  60  "  .i^v.  80  •a"  100 Depth (ft)  ;  •  -A r A-  120  140  160  Figure D3 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR THE 1:100 AND 1:475 EVENTS (MATHEWS ROAD SITE) (SOIL PROFILE BASED ON OLD SPT INFORMATION)  129  D4a Steveston (1:100 Year)  i  / V. G-  20  I ;  i Required (N1)60-Griffith Park Required (NljeO-Cal Tech Required (N1)60^Hughes Lake 12 Available (N1)60  f  v--  •  -O—o—O-^^v-.Q o-- :=fit=p=fO (i  a—i-e  :.*H.=a:^fira=;Depth -^:60 :^ &-^A--A^-g80 ' (ft)  100  40  120  D4b Steveston (1:475 Year) Requlred-(N_1)60=Qriffith. Park.. Required (N1J60-Cal Tech Required (N1)60-Hughes Lake 12 Available (N1)60 \  \,/ Hr ;  .. •'" -M-  20  -a. - A- - A- A — •-fl 40  I  I  -  60 Depth i  80  n.-.L  y -A-! ^ 100 \  120  Figure D4 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR THE 1:100 AND 1:475 EVENTS (STEVESTON SITE) (SOIL PROFILE BASED ON OLD SPT INFORMATION)  130  D5a Blundell (1:100 Year) Required (N1)60-Griffith Park Required (N1)60-Cal Tech Required (N1)6f>Hughes Lake 12 Available "(N 1)60  120  60 Depth (ft)  D5b Blundell (1:475 Year)  x  Required-(N1)60=Qriffith Park.. Required (N1J60-Cal Tech Required (N1)6r>Hughes Lake 12 Available "(Nil )60  60 Depth (ft)  120  Figure D5 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR THE 1:100 AND 1:475 EVENTS (BLUNDELL SITE) (SOIL PROFILE BASED ON OLD SPT INFORMATION)  131  APPENDIX E  132  E1a Tsawwassen (1:100 Year) CASE A - A/V= 1±0.2 Loma Prieta and San Fernando Records  ;  I  i,  i  i  i  Mean Moan 4- 1 Strl noviatinn  Avmailable (N1)60  I  * 20  40  60  80  100  Depth (ft)  E1b Tsawwassen (1:100 Year) CASEB -A/V= 1+0.5 Loma Prieta and San Fernando Records i  l  l  !  !  Mean Moan •+• 1 fitri Deviation  I  Avmailable (N1)60  ZZZZZZZZ-MZ-JZZZZZZ. •  \  „  1  1  1  •i^^J.  ^1  ;:::::!:;-•••••!  1 20  40  60  100  80  Depth (ft)  E1c Tsawwassen (1:100 Year) CASE C - A/V > 2 Loma Prieta and San Fernando Records Mean Mean + 1 Std. Deviation Available (N 1)60  100 Depth (ft)  Figure E1 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:100 EVENT, CASES A, B & C - ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  133  E2a Tsawwassen (1:100 Year) CASE D - Cat Tech and Griffith Park Records  !  |  ——-jy——}  .  i  l  l  !  !  ' A ,„..i  ~ ~ t  t 20  40  !  SPT (N1)60-Caljech A/V=1.48 SPT (N1)60-Cal Tech A/V= 1.66 SPT.(N1)60^Griffith.Pk..A/y.=0.88SPT (N1)60-Griffith Pk. A/V=1.17 Available (N1)60  }  | ;  S  I } I  |  ©••<  O i  "!  60  . \  \  80  100  Depth (ft)  E2b Tsawwassen (1:100 Year) CASEE - Loma Prieta and San Fernando Records  Mean £,_...  Mean + 1 Std. Deviation Available (N1)60  20  40  60  100  80,  Depth (ft)  E2c Tsawwassen (Mexico City 0.05g) CASEF- A/V =0.3 Mexico City Records  40 |  ;  ;  |  I  ;  ;  |  I  P*1  ;  |  1  j  ;  I  |  ; Mean  « £  Mean + 1 Std. Deviation 3  0  —I—;  I 0  j  i  j  i  I 20  i  ;  ;  i  !  i  i 40  ;  j  ;  i  i  .  Depth (ft)  i 60  ;  ;  i  i  Available (N1)60  i  I 80 < '  i  i  i  i 100  Figure E2 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:100 EVENT, CASES D, E & F - ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  134  E3a Tsawwassen (1:475 Year) CASE A - A/V= 1±0.2 Loma Prieta and San Fernando  40  1  Records  1 ! Mean Mean -H' 1 Std. Deviation Available (N1)60  1  o £ 30  .: O.  tn  J  ZZZb  .o ^ 20 o  ^:-;:::X  T  :::::::::::::  :::::  hOV"  : zigigE^  CO  """"--8  '  10 CO 20  40  60  Depth (ft)  80'  100  E3b Tsawwassen (1:475 Year) CASE B - AIV= 1+0.5 Loma Prieta and San Fernando  Records  40  1  : Mean  I £ 30 CO  §  Q.  20  i \  -  ,,,,.fh-  :::r==  z  I  ^^ii_„„.„„|o_::::::4  B—\ B I  1  B*  — o — -  Mean + 1 Std. Deviation Available (N1)60  •i  ^•--.^ ! Q"1 !—..]  10  CO  20  40  60  80  100  Depth (ft)  E3c Tsawwassen (1:475 Year) CASE C - A/V > 2 Loma Prieta and San Fernando  Records  40 •~ Mean Mean + 1 Std. Deviation Available (N1)60  I  0  £ 30 tn  1 20  j.  o  i  CO  10 IE 20  40  60  80  100  Depth (ft)  Figure E3 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:475 EVENT, CASES A, B & C - ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  135  E4a Tsawwassen (1:475 Year) CASE D - Cal Tech and Griffith Park Records  i  :=E  1  E  iijE=|EE|i=  2  1  \  l  l  ;  !  !  SPT (N1)60-CatTech A/V= 1.48 SPT (N1)60-Ca|Jech A/V= 1.66 SPT (N1)60-Griff]th Pk. A/V=0.88 SPT (N1)60-Griffjth Pk. A/V= 1.17. Available (N1)60 ;  i  20  i  i  40  i  i  i  60  100  80  Depth (ft)  E4b Tsawwassen (1:475 Year) CASE E - Loma Prieta and San Fernando Records  I  I  I  I  I  f  |  !.. ...,^  }  :;; ; v  ::  ^==^4^^^  !  !  i  0  = E 3 S E  I  Mean Mean + 1 Std. Deviation Available (N1)60  __j_,„..J_. |  i...............4 .^  I  i..::::e-"i  |  i  ••• -;wJQ-\ r;  f^.-0..j.-.:".:.:. j I  • 20  40  80  60  100  Depth (ft)  E4c Tsawwassen (Mexico City 0.05g) CASE F - A/V =0.3 Mexico City Records 40  i Mean Mean + 1 Std. Deviation Available (N1)60  £ 30  \  I 0  j  , a  i  i j  ) — — (  ::::::  n-  &-.-'- -', L  i 20  i  I  I  I 40  I  j  Depth (ft)  i  I 60  i  i  i  I 80  i  i  i  1 100  Figure E4 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:475 EVENT, CASES D, E & F - ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  /  136  E5a Ladner (1:100 Year) CASE A - A/V= 1±0.2 Loma Prieta and San Fernando Records 40.  X Mean  i  3 0 t  Mean + 1 Std. Deviation  I  Available (N1)60  CD 10 CO 10  20  30 Depth (ft)  50  40  60  E5b Ladner (1:100 Year) CASEB - A/V= 1±0.5 Loma Prieta and San Fernando Records Mean  —-© .Mean + 1 Std. Deviation Available (N 1)60  10  20  30 Depth (ft)  50  40  60  E5c Ladner (1:100 Year) CASE C - A/V > 2 Loma Prieta and San Fernando Records Mean  -©—Mean + 1 Std. Deviation Available (N1)60  10  20  30 Depth (ft)  40  50  60  Figure E5 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:100 EVENT, CASES A, B & C - ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  137  E6a Ladner (1:100 Year) CASED - Ca/ Tech and Griffith Park Records  i  1  i  I  SPT (N1)60-Cal Tech A/V== 1.48 SPT (N1)60-Caffech A/V== 1.66 SPT (N1)60-X3riff|th PR. A/V=a.88" SPT (N1)60-Griffith Pk. A/V-1.17 Available (N1)60 o  • & - i  r.r.f.r.r,.-,r,c-.  c>''-'T- '^P"-'4'"":  :  f •  :  :  7 : :  :  +•  :  :: ;  i 20  10  i  i  1  30 Depth (ft)  ,  40  60  50  E6b Ladner (1:100 Year) CASE E - Loma Prieta and San Fernando Records  i  I  I  R  i I  1 1 1 1 1 1 11 Mean Mean + 1 Std. Deviation Available (N1)60  !  II^z;!  - S  |  \ Q-i  |  i  |  !  {•• 1  10  siiiLiijirziiiiiiiiii-  20  30 Depth (ft)  40  60  50  E6c Ladner (0.05g) CASEF- A/V =0.3 Mexico City Records I  I  I  I  M  I  I I Mean  I  i  I  Mean + 1 Std. Deviation  Z  Available (N1)60  I ! ^ 8  10  20  30 Depth (ft)  40  50  60  Figure E6 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:100 EVENT, CASES D, E & F - ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  138  E7a Ladner (1:475 Year) CASE A - A/V= 1±0.2 Loma Prieta and San Fernando Records  1 Mean Mean + 1 Std. Deviation Available (N1)60 • •  |  |  [  |  Q i •  !  i  i  1 10  20  30 Depth (ft)  40  ...  1  i  i  50  60  E7b Ladner (1:475 Year) CASEB-A/V=  1 ±0.5 Loma Prieta and San Fernando Records !  I  I  !  Mean  --o-^  Mean + 1 Std. Deviation :: •-• I(N1)60 | | j Available  I (=  E} V i Qt \ • tf> i  D !  M  |  |  |  j  I  j |  |  j  j |  j  I i I i i I I 0  i i i i I i i i i I i i i iI  10  20  30 Depth (ft)  40  50  60  E7c Ladner (1:475 Year) CASE C - A/V > 2 Loma Prieta and San Fernando Records  I I I I I  Mean Mean + 1 Std. Deviation AvailablejN1)60  I I  i N ; ! :  [^"i~"4=""--  i  I  )  I  1 ; i i i I 0  10  «.-•( C;ilrrj==. =~4»H -1=4=-  i i ; j j i i i i i i i i i i i1  20  Depth 30 (ft)  40  50  60  Figure E7 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:475 EVENT, CASES A, B & C - ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  139  E8a Ladner (1:475 Year) CASED - Ca/ Tech and Griffith Park Records  40  ; i SPT (N1)60-CaLTech A/V= 1.48 SPT (N1)60-Ca£jech A/V=1.66 SPT (N1)60-Griffjth Pk. AA/=0.88 I SPT (N1)60-Griffjth Pkv A/V^4,17 Available (N1)60  I  ;  a  i.  Llll  •  I  0  I  0  i i i i  I  10  I i i i i :  j  20  i i  i i i i i i i i i i i i i i  i 30 Depth (ft)  40  50  I  60  E8b Ladner (1:475 Year) CASEE - Loma Prieta and San Fernando Records [  |  |  i  \  I  Q  #  I 0  i  i  i  i  I i 10  j  ! !  Mean Mean + 1 Std. Deviation Available (N1)60  i  j  I i 20  i  I  I I i i Depth 30 (ft)  i  i  1i 40  i  i  i  j  1i  i  i  50  i  1 60  E8c Ladner (Mexico City 0.05g) CASEF-A/V  = 0.3 Mexico City Records 1  !  Mean Mean + 1 Std. Deviation Available (N1)60  I  •  • -  0  \ 9 Si  i j—-6-r  i 1  10  20  30 Depth (ft)  40  50  60  Figure E8 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:475 EVENT, CASES D, E & F - ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  140  E9a Tsawwassen (1:100 Year) CASE A-A/V=  1 ±0.2 Loma Prieta and San Fernando Records  Mean  I  ....©...  I  I  I  Moan -l- 1 .Qtrl noviatinn A\/ailable (N1)60  *  - 8'  !T  X  ...... -  -820  40  60  s=:S-|  80  100  Depth (ft)  E9b Tsawwassen (1:100 Year) CASE B-A/V=  1±0.5 Loma Prieta and San Fernando Records  [ Mean  | i  i  Moan -I- 1 Rtri nouiatinn  I  A\/ailable (N1)60  .-.^*Z^  !'•'"'} t  "• 20  40  D " " i !  60  80  100  Depth (ft)  E9c Tsawwassen (1:100 Year) CASE C - A/V > 2 Loma Prieta and San Fernando Records  Mean Mean + 1 Std. Deviation Available (N1)60  20  40  60  80  100  Depth (ft)  Figure E9 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:100 EVENT, CASES A, B & C - 30' ABOVE THE ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  141  E10a Tsawwassen (1:100 Year) CASE D - Cat Tech and Griffith Park Records  40  1 SPT (N1)60-Caljech A/V= 1.48 SPT (N1)60-CalTech A/V= 1.66 SPT (N1)60-Griff]th Pk. A/V=0.88 SPT (N1)60-Griffith Pk.A/V=1.17 Available (N1)60  5 "••  o  20  CO  a.  =^=4^^+  *  1  10 1  8 20  .-A i  =6  40  60 Depth (ft)  i  -  I  i i  80  100  E10b Tsawwassen (1:100 Year) CASE E - Loma Prieta and San Fernando Records  Depth (ft)  E10c Tsawwassen (Mexico City 0.05g) CASE F - A/V =0.3 Mexico City Records  40  !  1  ! Mean  0  M p a n + 1 Stri  30  npviatinri  -  Av ailable (N1)60  1 n 20 o CO  10 D. CO  ::§::  20  '  40  ' 1 1  60  80  100  Depth (ft)  Figure E10 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:100 EVENT, CASES D, E & F - 30' ABOVE THE ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  142  E11 a Tsawwassen (1:475 Year) CASE A - A/V= 1±0.2 Loma Prieta and San Fernando Records  40  i  l l i Mean Mean + 1 Std. Deviation Available (N1)60  •  £ 30  l  I  -.vH .  I  ••  \ «  i i I i i I I IDepthi (ft) i I i  i  0  20  40  i  i  60  I i  i  i  80  I 100  E11 b Tsawwassen (1:475 Year) CASEB- A/V= 1±0.5 Loma Prieta and San Fernando Records  I  Mean  I  G  Mean + 1 Std. Deviation Available (N1)60  I  | -  \  b  •  20  40  60  80  100  Depth (ft)  E11 c Tsawwassen (1:475 Year) CASE C - A/V > 2 Loma Prieta and San Fernando Records  \  [  i Mean -— G—•  :  Y j : :  I:  Mean + 1 Std. Deviation Available (N1)60  t= 20  40  60  80  100  Depth (ft)  Figure E11 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:475 EVENT, CASES A, B & C - 30' ABOVE THE ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  143  E12a Tsawwassen (1:475 Year) CASED - Ca/ Tech and Griffith Park Records 40  i  I  SPT (N1)60-Cap"ech A/V=1.48 SPT (N1)60-Cafrech A/V= 1.66 SPT (N1)60-Griffjth Pk. A/V=0.88 SPT (N1)60-Griffjth Pk. A/V=1.17 Availably (ND60  o o 30  !  100  Depth (ft)  E12b Tsawwassen (1:475 Year) CASE E - Loma Prieta and San Fernando Records 40 Mean Mean + 1 Std. Deviation Available (N1)60 P  0 £ 30 tn  1  ^ 20 o  CO  •}"•:••"-O' |--' \ ::  T  -  CO 0  I i I i I i i I I i i i I i i i I i i i I  0  20  40  Depth (ft)  60  80  100  E12c Tsawwassen (Mexico City 0.05g) CASE F- A/V'» 0.3 Mexico City Records  I  I  Mean Mean + 1 Std. Deviation Available (N1)60  ; *  0  I i i i I i i i I iDepthi (ft) i I i i i I i i i 1  0  20  40  60  80  100  Figure E12 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:475 EVENT, CASES D, E & F - 30' ABOVE THE ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  144  E13a Ladner (1:100 Year) CASE A - AIV= 1+0.2 Loma Prieta and San Fernando Records  Mean Mean + 1 Std. Deviation Available (N1)60  10  20  30 Depth (ft)  40  50  60  E13b Ladner (1:100 Year) CASEB -A/V= 1±0.5 Loma Prieta and San Fernando Records  Mean Mean + 1 Std. Deviation Available (N1)60  10  20  30 Depth (ft)  40  50  60  E13c Ladner (1:100 Year) CASE C - A/V > 2 Loma Prieta and San Fernando Records  !  i  1 i Mean  !  !  1  MaoAvaila ri Xble1(N1)60 Ctrl Hov/iatirtn •  Q>i fc>! r  10  20  i i i :  30 Depth (ft)  40  50  60  Figure E13 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:100 EVENT, CASES A, B & C - 25' ABOVE THE ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  145  E14a Ladner (1:100 Year) CASE D - Ca/ Tech and Griffith Park Records  - •  :  -  SPT (N1)60-Caljech A/V= 1.48 SPT (N1)60-Ca(jech A/V= 1.66 SPT (N1)60-Griffjth Pk. A/V=0.88 SPT (N1)60-Griffjth Pk. A/V= 1.17Available (N1)60  \ j  1  - §:::::::::.§—'  1 ::::+::±  fiir:::::::::!:::]::::  = X = M - : ,  10  ••)  ;  20  30 Depth (ft)  40  j  [ 50  60  E14b Ladner (1:100 Year) CASE E - Loma Prieta and San Fernando Records X Mean Mean + 1 Std. Deviation Available (N1)60  10  20  30 Depth (ft)  40  50  60  E14c Ladner (Mexico City 0.05g) CASE F - A/V =0.3 Mexico City Records Mean Mean + 1 Std. Deviation Available (N1)60  10  20  30 Depth (ft)  40  50  60  Figure E14 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:100 EVENT, CASES D, E & F - 25' ABOVE THE ORIGINAL GROUND SURFAE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  146  E15a Ladner (1:475 Year) CASE A - A/V= 1±0.2 Loma Prieta and San Fernando Records  40  i  X  Mean  —u—  0  1 30 § o  Mean + 1 Std. Deviation Available (N1)60  20  CO 10  10  20  30 Depth (ft)  40  50  60  E15b Ladner (1:475 Year) CASE B - A/V= 1±0.5 Loma Prieta and San Fernando Records I  I  • t\i_Y=  :  I 0  i  Mean Mean + 1 Std. Deviation Available (N1)60  •  8  |:-i:i:::J:::: ,  ,  ,  ,  I , 10  ,  ,  ,  I , 20  ,  ,  , I ,  , I Depth 30 (ft)  ,  I • 40  ,  ,  ,  I 50  , , I. 60  E15c Ladner (1:475 Year) CASE C - A/V > 2 Loma Prieta and San Fernando Records  i  - •  =U  !  !  I  !  ! !  Mean Mean + 1 Std. Deviation Available (N1)60  I !  =  -  - it  i  1 ef—™  fB f  •oLl.JIZZ  I i i i i I i i i i I i j i i I i Ii i I i i i i I i i i i 1 0  10  20  30 Depth (ft)  40  50  60  Figure E15 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:475 EVENT, CASES A, B & C - 25' ABOVE THE ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  147  E16a Ladner (1:475 Year) CASED - Ca/ Tech and Griffith Park Records  I  i  ! I i ! !  SPT (N1)60-CaLJech AA/= 1.48 SPT (N1)60-Ca!Jech A/V= 1.66 SPT (N1)60-Griffjth Pk. A/V=0.88 I SPT (N1)60-GriffJtbPkvA/V=1.17Available_(N1)60  \ -I - >je i Qt™ l™ . -.j  ^.v.-.v • • •  10  20  30 Depth (ft)  40  50  60  E16b Ladner (1:475 Year) CASE E - Loma Prieta and San Fernando Records  |  I I  i ! T f I :\i  I  :  j ! ! I  !  Mean Mean + 1 Std. Deviation Available (N1)60  ! i  9y.—  •  !  •©—I—|—•—|—  10  20  30 Depth (ft)  40  50  60  E16c Ladner (Mexico City 0.05g) CASEF- A/V =0.3 Mexico City Records  --• I I  I fVi  I  Mean Mean + 1 Std. Deviation Availa ble Q(N1)60  I  •  10  20  30 Depth (ft)  40  50  60  Figure E16 COMPARISON OF IN-SITU AND EARTHQUAKE REQUIRED (N1)60 VALUES FOR A 1:475 EVENT, CASES D, E & F - 25' ABOVE THE ORIGINAL GROUND SURFACE (SOIL PROFILE BASED ON RECENT SCPT INFORMATION)  148  APPENDIX F  149  (lU) UOI}DA8|3  150  i — r  o o  in  o LO (m) U O I ; D A 9 G  Cxi  O  o  (ill) UOI}DA9|3  153  PERMISSION TO USE COPYRIGHTED MATERIAL Permission is hereby granted to  (Name of author of thesis)  3 $ttHT£TZ.  KEV/AV  (Title of thesis) EvALCJAftOflJ OF t&Srtic A^FSSAfeNT PJ&>rJ£VUB&  firmer  L-iaoeFAcrmti  University of British Columbia  AAJP  VetienMi^S  (Degree)  7b  ;  A"/ >f- ^ * -  (Year)  and to the National Library of Canada* to reproduce  (Figure/page numbers)  in (Title of article/book) IkL Irrffuew*. c£S'PT '^rnLP/liA.fP in  Sail LtQUt^tAtbr^ &t<^a*K£, Bva/ua/iams  (Name, issue number, and year of journal) €ar^-Jt^d^ £h  iZsuvck tedc  tyU  * UCB/££ZC  C  (Place, publisher and year of book) C^ofleje. <^ £nQ/* which will appear in this thesis.  (Signatures)  ;  Date Name of copyright holder Address  Return this form to:  Library - Special Collections 1956 Main Mall University of British Columbia Vancouver, B.C. Canada V6T 1Y3  Thesis Supervisor  •The National Library will lend or sell copies of the microfilm. All other publication rights are reserved.  T H E UNIVERSITY O F BRITISH C O L U M B I A 1956 Main Mall Vancouver, B.C., Canada V6T 1Y3  LIBRARY  Dear Sirs: As a graduate student at the University of British Columbia, I am preparing my thesis, which will be microfilmed by the National Library of Canada, and copies of the film will be lent or sold. May I have permission to use in my thesis, and for the National Library to microfilm, the excerpts from your publication(s) described on the back of this letter. I would be very grateful for your favourable consideration? of this request Would you please complete the form on the back of this letter, and return it to the address given on the form. Thank you very much.  Sincerely,  DE-11 (4/88)  PERMISSION T O U S E COPYRIGHTED  MATERIAL  Perrnission is hereby granted to (Name of author of thesis)  Gv/aLQATio/O OF ^SetSrAlC ASSE S.r MGKTT PROCElO UfigS PREDICT UQOEPACTION} Ajjfr P6Po£MftTio^S  (Title of thesis)  University of British Columbia (Degree)  H.HSc.  (year^  and to the National Library of Canada* to reproduce (Figure/page numbers)  in (Title of article/book) . SPT- Baled  /Ina/us'ts <J  (Name, issue number, and year of journal)  ^W^^^ecjr/VyS  ^tfc/t^  Z_k  U. 11>ocrb'ti 5e£D Mfjutark/- <£y#ipc&Lrfii: (Place, publisher and year of hf^kf^SiTscH ^JtMl^erS., /tfjQ which will appear in this thesis. (Signatures)  Date Name of copyright holder Address  Return this form to:  Library - Special Collections - Thesis Supervisor 1956 Main Mall University of British Columbia Vancouver, B.C. Canada V6T 1Y3  •The National Library will lend or sell copies of the microfilm. All other publication rights are reserved. _  T H E UNIVERSITY O F BRITISH C O L U M B I A 1956 Main Mall Vancouver, B.C., Canada V6T 1Y3  LIBRARY  Dear Sirs: As a graduate student at the University of British Columbia, I am preparing my thesis, which will be microfilmed by the National Library of Canada, and copies of the film will be lent or sold. May I have permission to use in my thesis, and for the National Library to microfilm, the excerpts from your publication(s) described on the back of this letter. I would be very grateful for your favourable consideration of this request Would you please complete the form on the back of this letter, and return it to the address given on the form. Thank you very much.  Sincerely,  DE-11 (4/88)  PERMISSION T O USE COPYRIGHTED MATERIAL Permission is hereby granted to (Name of author of thesis)  (Title of thesis)  B^Al-V  ftT(  ok)  OF . S ^ i S M t a  University of British Columbia (Degree) fi'^-Sc-  K^l/lNJ  -y  g\CHT£&  %  f\$S£S2>Mth)T  •  (Tear)  1 ^5  and to the National Library of Canada* to reproduce (Figure/page numbers)  in  (Title of article/book)  (Name, issue number, and year of Journal)  (Place, publisher and year of book) "TZtzAtiC/f  . T^to^W^  t^ner^y, Mine?  Resource. <>  which will appear in this thesis. (Signatures)  Date Name of copyright holder Address  Return this form to:  Library - Special Collections 1956 Main Mall University of British Columbia Vancouver, B.C. Canada V6T 1Y3  Thesis Supervisor  •The National Library will lend or sell copies of the microfilm. All other publication rights are reserved.  T H E UNIVERSITY O F BRITISH C O L U M B I A 1956 Main Mall Vancouver, B.C., Canada V6T 1Y3  LIBRARY  Dear Sirs: As a graduate student at the University of British Columbia, I am preparing my thesis, which will be microfilmed by the National Library of Canada, and copies of the film will be lent or sold. May I have permission to use in my thesis, and for the National Library to microfilm, the excerpts from your publication(s) described on the back of this letter. I would be very grateful for your favourable consideration of this request Would you please complete the form on the back of this letter, and return it to the address given on the form. Thank you very. much.  Sincerely,  DE-11 (4/88)  PERMISSION T O U S E COPYRIGHTED  MATERIAL  Permission is hereby granted to (Name of author of thesis) VCVi^l  (Title of thesis)  To  Pg^Ptc-r  L U A T t O A)  O F  SEISMIC  U<0u£FACTtOAj  ANf>  3T  Pf?OCgDU^£<7  A^E^SMC^T  Pg-^O|gMflTipAlS  University of British Columbia (Degree) H-A-Sa.  'gtCHTeg  (Tear)  1  ( ^ 7 .5  and to the National Library of Canada* to reproduce (Figure/page numbers)  in  (Title of article/book) /'So  ooo  Map  qzazf-Z  (Name, issue number, and year of journal)  (Place, publisher and year of book) 5oW&/S  arid  MtfV/AJ^  which will appear in this thesis. (Signatures)  Date Name of copyright holder Address  Return this form to:  Library - Special Collections - Thesis Supervisor 1956 Main Mall University of British Columbia Vancouver, B.C. Canada V6T 1Y3  •The National Library will lend or sell copies of the microfilm. All other publication rights are reserved.  T H E UNIVERSITY O F BRITISH C O L U M B I A 1956 Main Mall Vancouver, B.C., Canada V6T 1Y3  LIBRARY  Dear Sirs: As a graduate student at the University of British Columbia, I am preparing my thesis, which will be microfilmed by the National Library of Canada, and copies of the film will be lent or sold. May I have permission to use in my thesis, and for the National Library to microfilm, the excerpts from your publication(s) described on the back of this letter. I would be very grateful for your favourable consideration of this request Would you please complete the form on the back of this letter, and return it to the address given on the form. Thank you very much.  Sincerely,  DE-11 (4/88)  

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