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Energy based seismic design of a multi-storey hybrid building : timber-steel core walls Goertz, Caleb 2016

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Energy Based Seismic Design of aMulti-Storey Hybrid Building:Timber-Steel Core WallsbyCaleb GoertzB.A.Sc., The University of British Columbia, 2014A THESIS SUBMITTED IN PARTIAL FULFILLMENT OFTHE REQUIREMENTS FOR THE DEGREE OFMASTER OF APPLIED SCIENCEinTHE COLLEGE OF GRADUATE STUDIES(Civil Engineering)THE UNIVERSITY OF BRITISH COLUMBIA(Okanagan)April 2016c© Caleb Goertz, 2016        The undersigned certify that they have read, and recommend to the College of Graduate Studies for acceptance, a thesis entitled:    Energy Based Seismic Design of a Multi-Storey Hybrid Building: Timber Steel Core Walls  Submitted by                    Caleb Goertz                            in partial fulfillment of the requirements of   The degree of       Master of Applied Science                                           .  Dr. Solomon Tesfamariam, UBC Okanagan Supervisor, Professor (please print name and faculty/school above the line)  Dr. Thomas Tannert, UBC Vancouver Supervisory Committee Member, Professor (please print name and faculty/school in the line above)  Dr. Siegfried Steimer, UBC Vancouver Supervisory Committee Member, Professor (please print name and faculty/school in the line above)  Dr. Joshua Brinkerhoff, UBC Okanagan University Examiner, Professor (please print name and faculty/school in the line above)   External Examiner, Professor (please print name and university in the line above)   April 6, 2016 (Date submitted to Grad Studies)      AbstractThis thesis discusses a novel timber-steel core wall system for use inmulti-storey buildings in high seismic regions. This hybrid system combinesCross Laminated Timber (CLT) panels with steel plates and connections toprovide the required strength and ductility to core walled buildings. Thesystem is first derived from first principles and validated in SAP2000.In order to assess the feasibility of the system it is implemented in thedesign of a 7-storey building based off an already built concrete benchmarkbuilding. The design is carried out following the equivalent static forceprocedure (ESFP) outlined by the National Building Code of Canada forVancouver, BC. To evaluate the design bi-directional nonlinear time historyanalysis (NLTHA) is carried out on the building using a set of 10 groundmotions based on a conditional mean spectrum.To improve the applicability of the hybrid system an energy based design(EBD) methodology is proposed to design the timber-core walled building.The methodology is proposed as it does not rely on empirical formulas andforce modification factors to determine the final design of the structure.NLTHA is carried out on the proposed methodology using 10 ground motionsto evaluate the suitability of the method and the results are discussed andcompared to the ESFP results.iiiPrefaceThis thesis is based on the research work conducted in the School ofEngineering at The University of British Columbias Okanagan campus underthe supervision of Dr. Solomon Tesfamariam. All the literature review,mathematical calculations and simulations of this thesis are carried out bythe author. A list of my conference publications at The University of BritishColumbia are listed as follows.Conference Papers1. Goertz, C., Dickof, C., Ratzlaff, D., and Tesfamariam, S. 2016. Designand behaviour of a timber core-wall multi-storey hybrid building underseismic action. World Conference on Timber Engineering (WCTE),August 22-25, 2016, Vienna, Austria. (Accepted)2. Goertz, C., Mollaioli, F., and Tesfamariam, S. 2017. Energy-basedseismic design of a timber core-wall multi-storey hybrid building. WorldConference on Earthquake Engineering, January 9-13, 2017, Santiago,Chile. (Accepted)ivTable of ContentsAbstract . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . iiiPreface . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ivTable of Contents . . . . . . . . . . . . . . . . . . . . . . . . . . . vList of Tables . . . . . . . . . . . . . . . . . . . . . . . . . . . . . viiiList of Figures . . . . . . . . . . . . . . . . . . . . . . . . . . . . . ixList of Acronyms . . . . . . . . . . . . . . . . . . . . . . . . . . . xiiiNomenclature . . . . . . . . . . . . . . . . . . . . . . . . . . . . . xivAcknowledgements . . . . . . . . . . . . . . . . . . . . . . . . . .xviiChapter 1: Introduction . . . . . . . . . . . . . . . . . . . . . . . 11.1 Review of design methodologies . . . . . . . . . . . . . . . . . 31.1.1 Equivalent static force procedure . . . . . . . . . . . . 41.1.2 Performance based design . . . . . . . . . . . . . . . . 61.1.3 Research methodology . . . . . . . . . . . . . . . . . . 101.1.4 Objectives . . . . . . . . . . . . . . . . . . . . . . . . . 111.1.5 Organization of thesis . . . . . . . . . . . . . . . . . . 11Chapter 2: Energy based design . . . . . . . . . . . . . . . . . . 132.1 History . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 132.1.1 Design earthquake . . . . . . . . . . . . . . . . . . . . 15vTABLE OF CONTENTS2.1.2 Energy absorption and dissipation . . . . . . . . . . . 162.2 Design methodology . . . . . . . . . . . . . . . . . . . . . . . 19Chapter 3: Building systems . . . . . . . . . . . . . . . . . . . . 213.1 Cross laminated timber . . . . . . . . . . . . . . . . . . . . . 213.2 CLT buildings . . . . . . . . . . . . . . . . . . . . . . . . . . . 233.2.1 SOFIE projects . . . . . . . . . . . . . . . . . . . . . . 233.2.2 FPInnovations CLT testing . . . . . . . . . . . . . . . 253.2.3 University of Canterbury UFP post-tensioned timbercore walls . . . . . . . . . . . . . . . . . . . . . . . . . 283.3 Hybrid buildings . . . . . . . . . . . . . . . . . . . . . . . . . 293.3.1 SOM tall timber project . . . . . . . . . . . . . . . . . 293.3.2 mgb architecture + design and equilibrium consulting 313.3.3 UBC CLT infilled SMRF . . . . . . . . . . . . . . . . 333.4 Representative buildings . . . . . . . . . . . . . . . . . . . . . 35Chapter 4: Steel-timber core walls . . . . . . . . . . . . . . . . 384.1 Lateral resisting system . . . . . . . . . . . . . . . . . . . . . 384.1.1 T-stub connection . . . . . . . . . . . . . . . . . . . . 394.1.2 Brackets . . . . . . . . . . . . . . . . . . . . . . . . . . 414.2 Behaviour of timber-steel core walls . . . . . . . . . . . . . . 42Chapter 5: Multi-storey building design . . . . . . . . . . . . . 515.1 Gravity loading . . . . . . . . . . . . . . . . . . . . . . . . . . 525.1.1 Floor system . . . . . . . . . . . . . . . . . . . . . . . 535.1.2 Diaphragm . . . . . . . . . . . . . . . . . . . . . . . . 535.1.3 Beams and columns . . . . . . . . . . . . . . . . . . . 535.1.4 Connections . . . . . . . . . . . . . . . . . . . . . . . . 545.2 Modeling of lateral and gravity loading resisting systems . . . 595.2.1 CLT panels . . . . . . . . . . . . . . . . . . . . . . . . 605.2.2 Steel members . . . . . . . . . . . . . . . . . . . . . . 645.2.3 Connections . . . . . . . . . . . . . . . . . . . . . . . . 645.2.4 Frames . . . . . . . . . . . . . . . . . . . . . . . . . . 67viTABLE OF CONTENTSChapter 6: Seismic design and behaviour . . . . . . . . . . . . 686.1 Equivalent static force procedure . . . . . . . . . . . . . . . . 686.1.1 Ground motion selection . . . . . . . . . . . . . . . . . 716.1.2 NLTHA on the 3D timber-steel structure . . . . . . . 736.2 Energy based design . . . . . . . . . . . . . . . . . . . . . . . 776.3 Summary . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 93Chapter 7: Conclusions and future recommendations . . . . . 957.1 Summary of findings . . . . . . . . . . . . . . . . . . . . . . . 957.2 Conclusions . . . . . . . . . . . . . . . . . . . . . . . . . . . . 977.3 Future recommendations . . . . . . . . . . . . . . . . . . . . . 98Bibliography . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 100viiList of TablesTable 1.1 Performance levels with corresponding damage stateand drift limit ([Ghobarah, 2001] by permission frompublisher) . . . . . . . . . . . . . . . . . . . . . . . . . 7Table 5.1 Design loads and climatic information for the bench-mark building . . . . . . . . . . . . . . . . . . . . . . . 52Table 5.2 Steel gravity frame design details . . . . . . . . . . . . 54Table 5.3 K factors for CLT ( c© 2011, FPInnovations, by per-mission) . . . . . . . . . . . . . . . . . . . . . . . . . . 62Table 5.4 Orthotropic CLT properties . . . . . . . . . . . . . . . 64Table 5.5 Pivot hysteresis model parameters . . . . . . . . . . . 66Table 6.1 ESFP design details . . . . . . . . . . . . . . . . . . . 71Table 6.2 EBD process for steel-timber hybrid structure with1.5% target drift . . . . . . . . . . . . . . . . . . . . . 90Table 6.3 Final EBD design details . . . . . . . . . . . . . . . . . 91viiiList of FiguresFigure 1.1 3-D rendering of proposed timber-steel hybrid building 2Figure 1.2 ESFP force distribution . . . . . . . . . . . . . . . . . 4Figure 1.3 Definition of seismic performance factors ([Alam, Moniand Tesfamariam, 2012] by permission from publisher) 6Figure 1.4 Fundamentals of DDBD ([Priestley and Kowalsky,2000] by permission from publisher) . . . . . . . . . . 8Figure 1.5 Energy balance concept with force-displacement rela-tionship ([Choi et al., 2006] by permission from pub-lisher) . . . . . . . . . . . . . . . . . . . . . . . . . . . 10Figure 2.1 EBD flow chart . . . . . . . . . . . . . . . . . . . . . 20Figure 3.1 CLT panel layup ( c© 2011, FPInnovations, by permis-sion) . . . . . . . . . . . . . . . . . . . . . . . . . . . 21Figure 3.2 Rough CLT panel during manufacturing . . . . . . . 22Figure 3.3 7 storey CLT experimental test: a) building set up;b) plans for building ([Ceccotti et al., 2013] by per-mission from publisher) . . . . . . . . . . . . . . . . . 24Figure 3.4 7 storey CLT building connection failures: a) hold-down fastener failure; b) nail pull-out in bracket; c)compression perpendicular to grain in wood failure([Ceccotti et al., 2013] by permission from publisher) 25Figure 3.5 FPInnovations CLT wall test: a) test apparatus; b)panel during testing ( c© 2011, FPInnovations, by per-mission) . . . . . . . . . . . . . . . . . . . . . . . . . 26ixLIST OF FIGURESFigure 3.6 FPInnovations connection test: a) parallel-to-the-grain;b) perpendicular-to-the-grain ( c© 2015, Johannes Schnei-der, by permission) . . . . . . . . . . . . . . . . . . . 27Figure 3.7 FPInnovations CLT wall test showing angled connec-tion (left) and hold-down connection (right) ( c© 2011,FPInnovations, by permission) . . . . . . . . . . . . . 27Figure 3.8 CLT core walls with post tensioning and UFP connec-tions: a) 3D rendering of high seismic core option; b)UFP plate ( c© 2014, Andrew Dunbar, by permission) 29Figure 3.9 Coupled and single wall comparison ([Dunbar et al.,2014] by permission from publisher) . . . . . . . . . . 30Figure 3.10 SOM’s concrete jointed timber frame typical floorstructure ( c© 2013, Skidmore, Owings & Merrill LLP,by permission) . . . . . . . . . . . . . . . . . . . . . . 31Figure 3.11 FFTT system: a) Plan view; b) 3D rendering of coreexploded view ([Green and Karsh, 2012] by permis-sion through Creative Commons License) . . . . . . . 32Figure 3.12 FFTT system: a) Connection detail ([Green and Karsh,2012] by permission through Creative Commons Li-cense); b) Full embedment steel beam to CLT wallexperimental test ( c© 2013, Pooja Bhat, by permission) 33Figure 3.13 CLT infilled steel moment resisting frame: a) pro-posed method; b) elevation view of two bays infilledsix-storey structure ( c© 2013, Carla Dickof, by per-mission) . . . . . . . . . . . . . . . . . . . . . . . . . 34Figure 3.14 Wood Innovation and Design Centre ( c© 2016, EmaPeter, by permission) . . . . . . . . . . . . . . . . . . 35Figure 3.15 T3 Minneapolis ( c© 2016, Michael Green Architec-ture, by permission) . . . . . . . . . . . . . . . . . . . 36Figure 3.16 Brock Commons rendering ( c© 2016, Acton Ostry Ar-chitects, by permission) . . . . . . . . . . . . . . . . . 37Figure 4.1 3D rendering of bolted t-stub end plate connection . . 39xLIST OF FIGURESFigure 4.2 Structural details of the tested specimen ([Piluso andRizzano, 2008] by permission from publisher) . . . . . 40Figure 4.3 Experimental tension testing results for t-stub con-nection test 6 ([Piluso and Rizzano, 2008] by permis-sion from publisher) . . . . . . . . . . . . . . . . . . . 41Figure 4.4 Experimental cyclic tests on t-stub connection: a)yielding in welded connection; b) force-displacementcurve results for test B5 ([Piluso and Rizzano, 2008]by permission from publisher) . . . . . . . . . . . . . 41Figure 4.5 Core system deformation . . . . . . . . . . . . . . . . 43Figure 4.6 Free body diagram of timber-steel core system . . . . 44Figure 4.7 Multi-storey timber-steel core-wall system . . . . . . 47Figure 4.8 Storey drift of first principles and SAP2000 timber-steel core wall system . . . . . . . . . . . . . . . . . . 50Figure 5.1 Design building floor plan . . . . . . . . . . . . . . . . 51Figure 5.2 CLT to CLT wall or floor connection: a) double sur-face spline; b) half-lapped joint ( c© 2011, FPInnova-tions, by permission) . . . . . . . . . . . . . . . . . . 55Figure 5.3 CLT floor panel connection test setup ([Gavric et al.,2012] by permission from conference publishers) . . . 56Figure 5.4 Hysteresis loops of lap joint test ([Gavric et al., 2012]by permission from conference publishers) . . . . . . . 56Figure 5.5 CLT panel to steel beam connection: a) testing setup; b) results of monotonic and cyclic test ([Loss etal., 2015a] by permission from publisher) . . . . . . . 58Figure 5.6 High strength screws: a) 8, 10 and 12 mm VG CSKASSY Screw types with associated angled wedge washer;b) ASSY screw connecting steel plate to timber mem-ber ( c© 2014, MiTiCon Timber Connectors, by per-mission) . . . . . . . . . . . . . . . . . . . . . . . . . 59Figure 5.7 Model of proposed timber-steel core wall system . . . 60Figure 5.8 Orthogonal axes of a CLT panel . . . . . . . . . . . . 61xiLIST OF FIGURESFigure 5.9 Four node quadrilateral shell element ( c© 2013, Com-puters and Structures, by permission) . . . . . . . . . 63Figure 5.10 Pivot hysteresis model ( c© 2013, Computers and Struc-tures, by permission) . . . . . . . . . . . . . . . . . . 66Figure 5.11 T-stub connection: a) experimental result ([Pilusoand Rizzano, 2008] by permission from publisher); b)pivot model calibration in SAP2000 . . . . . . . . . . 67Figure 6.1 T-stub connections for 3D ESFP . . . . . . . . . . . . 70Figure 6.2 2% in 50 years uniform hazard spectrum . . . . . . . 72Figure 6.3 Pseudo acceleration response spectra for the ten setsof ground motions . . . . . . . . . . . . . . . . . . . . 73Figure 6.4 Bi-directional NLTHA on timber-steel hybrid structure 74Figure 6.5 Building drift results of NLTHA for 2% in 50 years inthe X-direction . . . . . . . . . . . . . . . . . . . . . . 75Figure 6.6 Building drift results of NLTHA for 2% in 50 years inthe Y-direction . . . . . . . . . . . . . . . . . . . . . . 75Figure 6.7 Interstorey drift results of NLTHA for 2% in 50 yearsin the X-direction . . . . . . . . . . . . . . . . . . . . 76Figure 6.8 Interstorey drift results of NLTHA for 2% in 50 yearsin the Y-direction . . . . . . . . . . . . . . . . . . . . 76Figure 6.9 EBD flow chart . . . . . . . . . . . . . . . . . . . . . 78Figure 6.10 Pseudo velocity response spectra . . . . . . . . . . . . 79Figure 6.11 MDOF to SDOF transformation . . . . . . . . . . . . 81Figure 6.12 Input energy for various target ductility . . . . . . . . 82Figure 6.13 Average acceleration for each trial period . . . . . . . 83Figure 6.14 Ratio of the equivalent velocity to pseudo velocity . . 84Figure 6.15 Ratio of inelastic to elastic input energy . . . . . . . . 85Figure 6.16 Hysteretic to input energy ratio . . . . . . . . . . . . 86Figure 6.17 Normalized axial distribution ratio . . . . . . . . . . . 88Figure 6.18 T-stub connections for 2D EBD . . . . . . . . . . . . 89Figure 6.19 Interstorey drift results for EBD . . . . . . . . . . . . 92Figure 6.20 Maximum displacement for EBD . . . . . . . . . . . . 92xiiList of AcronymsBRB Buckling restrained braceCISC Canadian institute of steel constructionCLT Cross laminated timberCMS Conditional mean spectrumDDBD Direct displacement based designEBD Energy based designESFP Equivalent static force procedureFFTT Finding the forest through the treesFNA Fast nonlinear analysisGSC Geological survey of CanadaIDA Incremental dynamic analysisMDOF Multi degree of freedomNBCC National building code of CanadaNLTHA Nonlinear time history analysisNRC National research centrePBSD Performance based seismic designRC Reinforced concreteSDOF Single degree of freedomUHS Uniform hazard spectrumxiiiNomenclatureA Cross sectional areaa Reduced restraint factorc Damping coefficientd3 Distance between the edge of the shear wall and the connectionE Elastic modulusE0 Elastic modulus in the parallel directionE90 Elastic modulus in the perpendicular directionEd Damping energyEe Elastic energyEh Hysteretic energyEi Input energyE∗i Modified input energyEi,elastic Elastic input energyEi,inelastic Inelastic input energyEk Kinetic energyEp Plastic energyE∗p Modified plastic energyE∗pM Modified plastic energy for MDOF systemEs Strain energyFA Force at connection AFB Force at connection BG Shear modulusG0 Shear modulus in the parallel directionG90 Shear modulus in the perpendicular directionH HeightxivNomenclatureHi Initial heighthn Total height of buildingI Moment of inertiaIE Importance factork StiffnessL LengthLe Effective lengthM1 First modal massMv Higher mode effect factorP Input forceR Seismic modification factorRd Ductility factorRo Overstrength factorS(Ta) Spectral acceleration at the period TaSv,g Maximum velocity relative to the groundT Period of the structureTa Specific period of the structureT1 Fundamental period of the structureu Displacementu˙ Velocityu¨ Accelerationu¨g Ground accelerationuT Total displacementuT,eq Total displacement for an equivalent SDOF systemuy Yield displacementuy,eq Yield displacement for an equivalent SDOF systemV Base shearVeq Equivalent velocityVy Yield base shearα Input energy modification factorα1 Pivot model factorα2 Pivot model factorβ Plastic energy modification factorxvNomenclatureβ1 Pivot model factorβ2 Pivot model factor∆ Total lateral deformation∆b Bending deformation∆N Relative storey deformation∆s Shear deformation∆T T-stub connection axial deformation∆1 Deformation due to anchorage∆2 Deformation due to shear and bendingγ Plastic energy MDOF modification factorµT Target displacement ductilityη Pivot model factorφt1 Fundamental mode shape vectors roof storey componentθi Relative storey rotation angleθN Storey rotation angleΓ1 Modal participation factorxviAcknowledgementsFirst of all, I would like to thank the faculty and my fellow studentsat UBC. In particular, I would like to thank my supervisor: Dr. SolomonTesfamariam, for his guidance and the opportunity to work on many researchprojects. I would also like to thank Dr. Fabrizio Mollaioli for his help andguidance on the EBD aspects of this thesis.I also thank Derek Ratzlaff and Carla Dickof for their input on thetimber-steel core system that was developed as part of a research project.The knowledge and expertise in structural design they shared is greatly ap-preciated. I would also like to thank family and friends for their support andencouraging me to take much needed research breaks. Most importantly, Iwould like to thank my parents for their endless support and encouragementthroughout my studies.xviiChapter 1IntroductionTall wood buildings have recently become popular in Canada with theintroduction of mass timber products (Gagnon and Pirvu, 2011; Green andKarsh, 2012). Mid-rise timber buildings are now becoming a viable optionfor developers and designers in British Columbia after the 2009 BC BuildingCode increased the height limitation on wood-frame structures from fourto six. Mass timber construction consists of smaller timber beams gluedtogether to create a large beam or panel capable of resisting higher loads.As western Canada is densely populated with forests, mass timber productsare an economic choice for builders in this region. Advantages of mass timberbuildings when designed and detailed properly include: a reduced carbonfootprint, constructability, aesthetics and construction time. Mass timberproducts are stiff and light and therefore can often result in brittle buildingswith high strength and low ductility. High seismic regions require a ductilestructure in order to reduce shear force demand on a building by allowingthe building to deform during the excitation.Recent studies on tall timber buildings (Ceccotti et al., 2013; Dunbar,Pampanin, and Buchanan, 2014; Gagnon and Pirvu, 2011; Green and Karsh,2012; SOM, 2013; Tesfamariam, Stiemer, Dickof, and Bezabeh, 2014) showthe benefits and possibility of mass timber tall buildings. However, onemain recurring problem with tall Cross Laminated Timber (CLT) buildingsis maximizing the potential of the CLT panels strength and stiffness throughthe connections. Conventional connection design studies have focused on an-gled brackets and hold-downs with self-tapping screws or nails connectingto the CLT. These types of connections have small contact areas and there-fore are subject to localized crushing of the CLT (Schneider, Karacabeyli,Popovski, Stiemer, and Tesfamariam, 2014). This localized crushing results1Chapter 1. Introductionin good plastic behaviour; however this sacrifices the potential future usabil-ity of the system. When these connections fail, typically the stresses in thepanel are not near the yielding point.This thesis proposes a CLT core system coupled with steel elements todissipate the earthquake load (Figure 1.1).Figure 1.1: 3-D rendering of proposed timber-steel hybrid buildingThe proposed system utilizes CLT panels with steel plates and ductilesteel connectors that transfer the seismic load from the CLT floor to the21.1. Review of design methodologiesfoundation. Steel t-stub connections provide the ductility to the system byplastically deforming at the base of each floor. The plates were designedto transfer the shear forces from the CLT panel to the t-stub connections.Moreover, steel brackets transfer the shear forces from the diaphragm to thewalls. The proposed lateral system is supported by a steel gravity systemthat efficiently carries loads from the roof down to the concrete foundation.The aim of the proposed system was to maximize the strength and stiffnessof the CLT panels within the core walls by using steel plates that run theheight of the panels and are continuously connected to the panel.1.1 Review of design methodologiesPast earthquakes have shown the fault in current seismic design philos-ophy through unexpected damages, economic loss, and costly repairs of ap-propriately code detailed and designed buildings (Ghobarah, 2001; Pang andRosowsky, 2009). These factors need to be addressed so that future earth-quakes do not result in similar outcomes; however, current design method-ologies do not account for these factors due to the simplified code regula-tions. Current earthquake design philosophy in North America recommendsan equivalent static force procedure (ESFP). The ESFP relies on spectralacceleration response spectrums to calculate the base shear required by thestructure to resist. This spectral acceleration response spectrum is predictedbased on uniform hazard spectrums (UHS) developed through attenuationrelationships. The attenuation relationships give acceleration values for var-ious periods accounting for source to site distance, focal depth, magnitude,and shear wave average velocity (Boore, Joyner, and Fumal, 1997). Toaddress the current seismic design problems research has focused on deter-mining the realistic performance of the structures to estimate the loss anddamages an earthquake event would cause. This method was termed per-formance based seismic design (PBSD). The following sections give briefoverviews of the current seismic design methodology and proposed PBSDmethods.31.1. Review of design methodologies1.1.1 Equivalent static force procedureCurrently in Canada, and most parts of the world, the building coderecommends the ESFP. The method is relatively simple and can be com-pleted by hand without the use of finite element software. For this reason,designers have become accustomed to the method and enjoy the relativesimplicity to the seismic design process. The building is designed to resist aforce determined through ESFP known as the base shear. This base shearis distributed along the height of the building based on the mass of the floorwith an inverted triangular distribution as shown in 1.2.V Figure 1.2: ESFP force distributionThe basic seismic design principles have not changed much since theintroduction of the first National Building Code of Canada (NBCC) in 1941.Since then, relatively minor changes have been made to this code. Mostnotably, the 2005 edition of the NBCC saw the most change by introducingductility and overstrength factors to replace the force modification factor(R). These changes resulted in the formation of a new base shear equation:V =S(Ta)MvIeWRdRo(1.1)where V is the base shear to be applied on the building; S(Ta) is thedesign spectral response acceleration at the fundamental period Ta; Mv ac-counts for the higher mode effects on the base shear; Ie is the importance41.1. Review of design methodologiesfactor; W is the weight of the structure; Rd is the ductility factor; and Rois the overstrength factor (National Research Council (NRC), 2010). Upperand lower bounds on the seismic base shear value were also proposed; theminimum base shear:V ≥ S(2.0)MvIeWRdRo(1.2)and the maximum base shear:V ≤ 23S(0.2)IeWRdRo(1.3)Both the upper and lower bound of the seismic base shear were intro-duced to ensure buildings in low seismic regions still have some lateral re-sistance and that buildings were not overdesigned based on the proposedlocation. The ESFP is well defined and has been thoroughly studied as it isthe current standard in most countries. More details on ESFP can be foundin the 2010 edition of the NBCC (National Research Council (NRC), 2010).The ductility and overstrength factors relating to the structural systemcan be found in part 4 of the NBCC. These factors account for the plas-tic behaviour and designed overstrength in the system as shown in Figure1.3. Unfortunately, the ductility and overstrength factors remain constantfor a structural system regardless of the height of the structure. Studieshave shown that the ductility of a system changes as the building height in-creases (Dickof, Stiemer, Bezabeh, and Tesfamariam, 2014; Tesfamariam etal., 2015; Zhang, Fairhurst, and Tannert, 2015). Therefore, designing withthese factors can result in over-designed and potentially inefficient buildings.The discussed force modification factors vary based on the structure ma-terial and type. When designing complicated systems containing two struc-tural systems or two different materials the more conservative modificationfactor governs the design base shear (National Research Council (NRC),2010). This, however, can lead to over designed and costly buildings. With51.1. Review of design methodologiesFigure 1.3: Definition of seismic performance factors ([Alam, Moni andTesfamariam, 2012] by permission from publisher)the current knowledge of earthquake events and computational technologyavailable, a much more rigorous design method would show the performanceof these complicated structures.1.1.2 Performance based designMuch research lately has been in new performance based methodolo-gies including direct displacement based design (DDBD) and energy baseddesign (EBD). PBSD emerged to provide design guidelines that enable thedesigner to meet a predetermined performance level for a particular location(Ghobarah, 2001). The building codes addressed life safety in seismic designcontrolling the damage for small and moderate earthquakes and preventingcollapse in major earthquakes. Performance of the building is defined bythe displacement and interstorey drift that the structure would see duringthe earthquake events. PBSD allows the designer to determine the designreliability to achieve the design objectives. Several fundamental reports con-tributed to the development of PBSD (ATC, 1996; FEMA, 1997; SEAOCVision 2000 Committee, 1995). Ghobarah (2001) summarized the perfor-61.1. Review of design methodologiesmance levels corresponding to damage state and drift limits as shown inTable 1.1.Table 1.1: Performance levels with corresponding damage state and driftlimit ([Ghobarah, 2001] by permission from publisher)Displacement based designDDBD does not rely on empirical formulas and acceleration responsespectrum but rather relies on design displacement, effective mass, and theeffective height of the building. The first true DDBD method was introducedin 1993 by Priestley (Priestley, 1993). The fundamental steps of DDBD canbe seen in Figure 1.4. The method involves estimating the deformation ofa single degree of freedom (SDOF) system that represents a multi degreeof freedom (MDOF) systems first fundamental period (Chopra and Goel,2001) (Figure 1.4a). After the equivalent SDOF is determined the stiffnessand damping properties are examined (Figure 1.4b,c). The displacementof the equivalent SDOF system is then determined based off the designdisplacement response spectra relative to the earthquake hazard level andfundamental period (Figure 1.4d).Energy based designResearch in EBD has not had the attention of DDBD yet now is gain-ing in popularity because of the methods reliance on the velocity spectrumand incorporates duration of earthquake hazard. Although structural dam-71.1. Review of design methodologiesFigure 1.4: Fundamentals of DDBD ([Priestley and Kowalsky, 2000] bypermission from publisher)age from an earthquake to a structure is caused solely by deflections, en-ergy based methods allow for the assessment of the energy dissipated andabsorbed from the earthquake event. This assessment is helpful in design-ing the building by balancing the energy seen by the structure (Mollaioli,Bruno, Decanini, and Saragoni, 2011). In order to understand the demandsan earthquake puts on a structure it is important to consider that durationeffects as energy is a cumulative measure of ground shaking. By consideringduration the fault type can be included in the determination of seismic forcesincluding near-fault ground motions. Energy based design involves two keyaspects, the first being establishing design earthquakes and the second beingdetermining the actual absorption and dissipation capacity of the structure(Decanini and Mollaioli, 1998). The concept of energy can be derived from81.1. Review of design methodologiesthe equation of motion:mu¨+ cu˙+ ku = −mu¨g (1.4)where m is the mass of the system, c is the viscous damping coefficient,k is the stiffness, u is the displacement and ug is the ground acceleration.By integrating each term with respect to the relative displacement, u, in theequation of motion the energy equation can be derived as:∫mu¨du+∫cu˙du+∫kudu = −∫mu¨gdu (1.5)The three terms on the left side of Equation 1.5 are related to the struc-tural characteristics and represent the stored kinetic energy (k), dissipatedenergy through damping (Ed), and absorbed energy (Ea), respectively. Theabsorbed energy; however, can be further separated into strain energy (Es)and hysteretic energy dissipation (Eh). Strain energy represents the recover-able energy that the structure can withstand whereas the hysteretic energydissipation represents the irrecoverable hysteretic energy which causes thedamage to the structure. These energies, when summed equate to the to-tal input energy subjected by the earthquake (Ei). Therefore, the energyequation can be simplified as:Ek + Ed + Es + Eh = Ei (1.6)Researchers have developed various design methodologies using the en-ergy principles described above(Choi and Kim, 2005, 2009; Ghosh, Adam,and Das, 2009; Leelataviwat, Goel, and Stojadinovi, 2002). The fundamen-tal concept in EBD is the energy-balance concept as shown in Figure 1.5.This figure shows that the elastic and plastic energy when summed shouldequate to the input energy of an equivalent elastic system at the maximum91.1. Review of design methodologiestarget displacement (Choi, Kim, and Chung, 2006). These studies displayedthat the target performance was met when the designs were subjected toanalytical design earthquakes.Figure 1.5: Energy balance concept with force-displacement relationship([Choi et al., 2006] by permission from publisher)1.1.3 Research methodologyDesign and analysis were carried out on a multi-storey building basedon an already built concrete building that utilized concrete core walls. Thesteel column, beams and timber flooring were sized according to the gravityloads as designed for the benchmark building. After the gravity system wasdefined the timber-steel core walls were designed first using the conventionalESFP method and validated using bi-directional 3D nonlinear time historyanalysis (NLTHA). NLTHA consists of subjecting the analytical inelasticmodel of the building to a scaled earthquake ground motion. Bi-directionalground motions were used to better understand the behaviour of the build-ing under a realistic earthquake event. Results of the NLTHA predict thedisplacement and interstorey drift of the building due to the ground motion.The building was then designed using an EBD methodology and compared101.1. Review of design methodologieswith the ESFP results. The results are discussed in this thesis for the pro-posed timber-steel hybrid mid-rise residential building.To study the ESFP and EBD designed buildings NLTHA was carriedout using the finite element software SAP2000 (CSI, 2013) .The softwareallows designers to analyze a buildings behaviour using the input materi-als, elements and loading. Powerful finite element programs are becomingnecessary for new and complex building designs as they can validate thebuildings performance and ensure a safe design.1.1.4 ObjectivesThis thesis has multiple objectives including:− Designing a timber-steel core wall system suitable for high seismicregions that fully utilizes the strength and stiffness potential of CLTpanels.− Designing the proposed structure following the NBCC recommendedESFP for seismic analysis and analyzing the designed structures be-haviour with bi-directional NLTHA.− Designing the proposed structure using an EBD methodology and com-paring the results to the ESFP design.1.1.5 Organization of thesisThis thesis is organized into 7 chapters. Chapter 2 discusses energybased design including the history and the method used in this thesis todesign the proposed structure.Chapter 3 discusses building systems. First, the relatively new masstimber product CLT is discussed in detail. Next, relevant CLT buildingsand CLT hybrid buildings past studies are discussed. .Chapter 4 explains the proposed timber-steel core wall system. Eachcomponent of the system is defined and the lateral behaviour is defined usingfirst principles.111.1. Review of design methodologiesChapter 5 outlines the multi-storey building design carried out for thetimber-core wall system. The designed gravity load bearing system is dis-cussed and the resulting plans are shown. SAP2000 modeling details arealso discussed for all parts of the building including: CLT areas, links, steelframes, loading and releases.Chapter 6 includes the design of the lateral system of the building usingthe steel-timber core walls. The ESFP design and results are first discussedfollowed by the EBD design and results. Finally a discussion of the resultsis presented for both methods.Chapter 7 presents final conclusions of both the ESFP and EBD designmethods for the proposed steel-timber hybrid system and future recommen-dations for related research.12Chapter 2Energy based designEnergy based design is discussed in detail in the following sections includ-ing the history of energy based design, the energy approach for determiningthe seismic energy input to a structure and finally the structural design todissipate the energy appropriately. The main concepts of EBD can be foundin Chapter 1.2.1 HistoryWhile much focus in performance based design methods was on dis-placement based design, a few researchers studied performance based designthrough an energy approach. Housner first proposed the idea of energybased design in 1956 when he observed that the energy input into the struc-ture from an earthquake is dissipated partially through damping while therest is stored through kinetic energy (Housner, 1956). Therefore, he sug-gested that buildings should be designed to absorb energy plastically duringextreme earthquake events. Furthermore, he proposed an equation for theinput energy on a structure based on the mass and velocity response spec-trum:Ei =12mS2v g(2.1)where Ei is the input energy; m is the mass; and Sv,g is the maximumvelocity relative to the ground. The concepts suggested in Housners paperare still currently the main fundamental seismic design principles. Moreover,designing buildings to remain elastic during the maximum earthquake event132.1. Historyis not economical and is politically too extreme as the building would beover-designed and expensive (Estes, 2003). Therefore, buildings are nowdesigned to withstand a maximum earthquake event by entering the plasticzone but still preventing collapse and most importantly maintaining thesafety of the occupants.Little research followed Housner’s findings on energy based design un-til the mid-1970’s (Kato and Akiyama, 1975; Soni, Krishna, and Chandra,1977). Kato and Akiyama (1975) studied the link between the energy the-ory and earthquake damage in structures accounting for inelastic behaviourusing the ductility factor concept. The concept of energy based design is dis-cussed by the authors but not explicitly used. Soon after, researchers Soni,Krishna and Chandra proposed a design methodology in 1977 based on themechanics of energy absorption governed by force and deflection criteria.Kato and Akiyama (1982) followed up their work from 1975 and de-signed a steel structure using energy based design principles. If, the energyabsorption in the structure was greater than the energy input into the struc-ture, the structure was deemed safe. This work triggered the investigationinto energy based design concepts (Akiyama, 1985; Mccabe and Hall, 1989;Tembulkar and Nau, 1987; Uang and Bertero, 1988a, 1988b; Zahrah andHall, 1984). Most significantly, Akiyama (1985) wrote a book on the earth-quake resistant limit state design of buildings in which he focused on theeffect hysteretic damping has on a structure with regards to energy. Uangand Bertero published two papers in 1988 on the matter of energy baseddesign. The first paper stressed the importance of duration in the groundmotion and the recommended method in constructing input and hystereticenergy spectra which corresponds to all types of ground motions that couldoccur on the site; moreover, the second paper focused more on the use ofenergy in seismic design. Researchers McCabe and Hall pointed out that thepresent design philosophy is equilibrium based rather than deformation orcompatibility-based (Mccabe and Hall, 1989). Their paper follows with dis-cussion on the influence various factors have on structures such as frequency,yield resistance, duration of excitation and energy absorption.Further studies (Fajfar, Vidic, and Fischinger, 1989, 1990; Leger and142.1. HistoryDussault, 1992; Uang and Bertero, 1990) on energy based design in thelate 80’s and early 90’s confirmed it was the more realistic seismic designprocess. Soon after, two papers by Bruneau and Wang studied the responseof single-degree-of-freedom structures to earthquakes with an energy basedapproach (Bruneau and Wang, 1996a, 1996b). After Bruneau and Wang’spapers, the interest in energy based design principles increased significantlyand a seismic design method was proposed in 1999 (Leelataviwat, Goel, andStojadinovic, 1999).2.1.1 Design earthquakeThroughout the last 20 years researchers have studied earthquake datawith aims of developing an accurate earthquake input energy spectra. Theproposed earthquake input energy spectra’s were first formulated using afew parameter such as magnitude, source to site distance, site class andductility factor (Chou and Uang, 2000). However, through extensive studiesthe input energy spectra was developed using many more factors such asduration, fault type, depth and peak ground velocity (Cheng, Lucchini, andMollaioli, 2014).Extensive studies on energy based design spectrum started in the late1990’s and early 2000’s (Benavent-Climent, Pujades, and Lopez-Almansa,2002; Cruz A. and Lpez, 2000; Decanini and Mollaioli, 1998; Decanini andMollaioli, 2001; Manfredi, 2001; Riddell and Garcia, 2001; Trifunac, Hao,and Todorovska, 2001). Decanini and Mollaioli (1998) introduced the pro-cedure to determine elastic design earthquake input spectra by showing theimportance of various factors on the input with particular interest in theproximity to source factor. Further studies (Chou and Uang, 2000) devel-oped attenuation relationships that accounted for magnitude, source to sitedistance, site class and ductility factor. Furthermore, Manfredi (2001) dis-cussed the importance of hysteretic input energy by predicting the energyspectra from an earthquake.Similar studies (Ordaz, Huerta, and Reinoso, 2003; Trifunac, 2005, 2008)investigated the energy input to a structure. Ordaz et al. (2003) studied the152.1. Historyeffect of magnitude and focal distance on the input while Trifunac (2005,2008) studied the effect of power incident wave-pulses in comparison withthe capacity of a structure. Studies by Kalkan and Kunnath (Kalkan andKunnath, 2007, 2008) brought great knowledge to the field of energy demanddue to earthquakes. Their 2007 study suggested a new way in measuringthe severity of ground motions deemed as the peak-to-peak energy demand.This concept considered the severity of the earthquake through effectivecyclic energy demand. Moreover, Kalkan and Kunnath’s study in 2008 ex-amined the effects of near fault ground motions from a past study (Hall,Heaton, Halling, and Wald, 1995) on a variety of structures. The authorsfound that the near-fault accelerograms should be based on the shape andperiod of the dominant pulse in the corresponding record, as well as theproperties of the structural system, in order to appropriately define the en-ergy input to a structure. Finally, most recently researchers Cheng, Lucchiniand Mollaioli proposed new ground-motion prediction equations with regardto the elastic input energy spectra (Cheng et al., 2014). The comprehen-sive study developed these prediction equations using empirical regressionon mixed-effect models on a variety of strong-motion ground records.2.1.2 Energy absorption and dissipationThe structural designer is tasked with distributing this input or hys-teretic energy on the structure to provide an efficient but most importantlysafe building. Doing so requires rigorous research on the energy based designprinciples such as total energy, hysteresis energy and damping energy in therelevant structural system.In the early 2000’s several studies (Akba, Shen, and Hao, 2001; Akiyama,2002; Choi and Shen, 2001; Chou and Uang, 2003; Leelataviwat et al., 2002;Wong and Wang, 2001) investigated various structural systems to balancethe energy formula for input energy. Akbas et al. (2001) studied the energydemand for total energy, hysteresis energy, and damping energy for steel mo-ment resisting frames (SMRF) which were represented by their fundamentalperiod. The study distributed the energy to individual members within the162.1. Historystructure and found that the cumulative plastic rotation capacity was vitalin EBD of SMRFs. Similarly, Leelataviwat et al. (2002) designed a six storeySMRF by adding the distribution of base shear on the height of the structureand the plastic yield mechanism to the energy balance concept. The EBDmethodology was furthered when studies focused on the force-deformation(P−∆) effect due to earthquakes and gravity loading; therefore, consideringthe total input energy (Akiyama, 2002). Soon after, Chou and Uang (2003)presented a procedure that converted multi degree-of-freedom structures tosingle degree-of-freedom structures in order to predict the seismic energydemand at each floor without nonlinear time history analysis.A significant contribution to the evolution of EBD is due to the reportpublished by the National Institute of Standards and Technology: Distribu-tion of Earthquake Input Energy in Structures (Khashaee, Mohraz, Sadek,Lew, and Gross, 2003). The detailed report focuses both on the influenceof ground motion characteristics as well as the distribution of input energyinto a structure. Similar in detail, Estes thesis from the University of South-ern California provided a thorough review on EBD and proposed a designmethodology based on plastic rotation for steel moment resisting frames(Estes, 2003).Further research on the energy based design methodology was explored inthe mid 2000’s (Chang and Kawakami, 2006; Choi and Kim, 2005; Hernndez-Montes, Kwon, and Aschheim, 2004; Parulekar, Vaity, Reddy, Vaze, andKushwaha, 2004; Smyth and Gjelsvik, 2006). These studies furthered theenergy based design method by proposing new methods analytically as wellas experimentally. The study by Choi and Kim (2005) proposed a designmethodology for buckling-restrained braces and included a full detailed de-sign. The proposed methodology determined the input energy to the equiva-lent single degree of freedom (SDOF) system through time history analysis ofground motions. This equivalent SDOF system was then converted back toa multi degree of freedom (MDOF) system where the energy was distributedto each storey based on the hysteretic energy demand of the building.Recent research of designing structures to withstand earthquake forcesusing energy based concepts have displayed how current design methodolo-172.1. Historygies fall short in dissipating energy throughout a structure (Jara, Miranda,and Ayala, 2007; Lucchini, Mollaioli, and Monti, 2011; Mollaioli and Bruno,2008; Mollaioli et al., 2011). Jara et al. (2007) studied energy dissipatingdevices in SDOF structures and found the inelastic response of the sys-tems with special attention to soft-soil sites. Further research (Mollaioliand Bruno, 2008) has shown that soil-site conditions along with fundamen-tal periods and ductility ratios have a large influence on the displacementand drift ratios of structures subjected to ground motions.Energy based design has focused on steel structure design most sig-nificantly with recent research focusing on steel moment resisting frames(SMRF) and braced frames (Choi and Kim, 2009; Leelataviwat, Saewon,and Goel, 2009). Leelataviwat et al. (2009) proposed a new nonlinear staticprocedure based on energy concepts, developing and designing a three storeySMRF to validate their method. The design validated that MDOF systemscan be designed based off the energy balance concept. Following up theirstudy in 2005, researchers Choi and Kim studied the proposed energy baseddesign methodology on buckling-restrained braced frames (Choi and Kim,2009). The researchers found that the methodology underestimated thestrength demand of structures and recommended that structures be checkedfor strength after designed based on energy as strength-based design wouldrequire a displacement check.Several researchers have studied reinforced concrete (RC) design usingenergy based design (Acun, 2010; Benavent-Climent and Zahran, 2010; Caoand Friswell, 2009; Lestuzzi and Bachmann, 2006; Lucchini, Mollaioli, andBazzurro, 2014; Terapathana, 2012). Significant works to the knowledge ofenergy based design were theses completed by Acun (2010) and Terapathana(2012). The theses both studied energy based design of RC where Terap-athana (2012) focused on RC structures and Acun (2010) focused on RCcolumns. Using incremental dynamic analysis (IDA) researchers Benavent-Climent and Zahran (2010) found that at the collapse of RC structures mostof the hysteretic energy is concentrated in one single storey. Moreover, ex-periments confirmed that 90% of the energy dissipation was from plasticdeformation (Lestuzzi and Bachmann, 2006).182.2. Design methodologyResearch on energy based design only considered steel and concrete struc-tures until recently a team of researchers developed a new system for timberplatform frame buildings with energy dissipaters (Lopez-Almansa, Segues,and Cantalapiedra, 2014). The researchers studied the proposed system withenergy based design principles and finished the study by designing a build-ing with the proposed solution. Furthermore, the researchers discussed thethree major advantages of energy based design over other design methods:first, the energy input into the structure is almost totally uncoupled withthe dissipation capacity; second, the input energy is a stable quantity that isonly affected by the mass and natural period of the structure in the mediumto long period range; and third, cumulative damage can be appropriatelyaddressed (Lopez-Almansa et al., 2014).2.2 Design methodologyThe EBD methodology used in this thesis follows the research methodproposed by Choi et al. (2006). The research method was followed due tothe methods reliance on energy input and modified energy balance conceptas shown in Chapter 1. The researchers design methodology was designedfor a buckling-restrained brace (BRB) structure. Therefore, their designmethodology was modified for use with this thesis’ proposed steel-timbercore wall system. Figure 2.1 depicts a flow chart for designing BRB framedstructures according to Choi et al., 2006.192.2. Design methodology1. Determination of yield displacement 2. Estimation of input energy 3. Estimation of the yield base shear 4. Estimation of the total plastic energy 5. Story-wise distribution of plastic energy 6. Determination of the cross-sectional area of the buckling-restrained braces in each storey Figure 2.1: EBD flow chartThe proposed method adopted from Choi et al. (2006) can be found inChapter 6 of this thesis. The details of the proposed method are discussedand the method is carried out on the proposed steel-timber core-wall hybridstructure.20Chapter 3Building systems3.1 Cross laminated timberCLT is an orthotropic panel made up from dimensioned lumber andadhesive. In British Columbia, the dimensioned lumber is spruce-pine-firnumber two or better and the adhesive changes based on the type of pressof the CLT manufacturing plant, open time and price. The dimensionedlumber is placed edge to edge to make up the bottom layer of the panel.On top of this first layer of lumber a second layer is oriented 90 degrees tothe boards below (Figure 3.1). This process is repeated until the desiredamount of layers is achieved with a minimum of 3 layers. The result is astrong and stiff panel capable of resisting large loads with a variety of designapplications. The panels were originally designed to compete with two-wayconcrete slabs; however, they also show great behaviour as walls, shear wallsand roofs.Figure 3.1: CLT panel layup ( c© 2011, FPInnovations, by permission)213.1. Cross laminated timberBetween each layer of dimensioned lumber an adhesive is applied tothe boards securing the crosswise boards together. Once the desired panelthickness is reached the panel is placed in a press until the adhesive hascured. Out of the press, the panel is in rough shape and requires sandingand planing (if required) as shown in Figure 3.2. The resulting product isan aesthetically pleasing pre-fabricated panel ready to be placed on site.Figure 3.2: Rough CLT panel during manufacturingThere are numerous reasons to use CLT, the most noteworthy being thesustainability of the product when compared with other building products(e.g. concrete and steel). Timber is a carbon sequestering material: ab-sorbing carbon dioxide from the environment and releasing it back to theenvironment once burned. However, by burning the timber product at theend of its life cycle heat can be provided for buildings or systems whereanother source of energy would be required. Life cycle assessment studieson different building material options show that CLT was superior in termsof carbon footprint to steel and concrete options (Gagnon and Pirvu, 2011).Even though CLT has been rigorously studied and the manufacturershave met the strict criteria for the product, it is still not common in designchoices for new construction. Many factors affect the slow adoption of CLTin North America including: public perception on fire safety, lack of manu-facturers, and high cost. Fire testing on CLT at FPInnovations in Quebec,however, has showed acceptable performance of CLT due to the slow char-223.2. CLT buildingsring rate (Gagnon and Pirvu, 2011). As the current cost of CLT productionis high, projects on a limited budget will quickly eliminate CLT options inthe early stages of conceptual design.3.2 CLT buildingsSeveral studies have explored the behaviour of common CLT buildingswith steel connectors. The relevant studies are presented in the followingsections.3.2.1 SOFIE projectsNumerous studies on X-lam (CLT) have been carried out under theSOFIE project to investigate its material behaviour through a variety oftests including cyclic tests on wall panels, pseudo-dynamic tests and full-scale tests on shake tables (Ceccotti et al., 2013). Although the studies havenot been hybrid buildings, the works have largely contributed to the CLTwork to date. Moreover, most of the design effort for hybrid CLT build-ings is in connection detail. Experimental tests were conducted on X-lamwalls under lateral force with cyclic and pseudo-dynamic tests and the re-sults were studied by various researchers (Ceccotti, Lauriola, Pinna, andSandhaas, 2006; Gavric, Fragiacomo, and Ceccotti, 2012, 2015).Observations from the failures in the timber system displayed that thetimber panels acted almost completely rigid (Ceccotti et al., 2006). Conse-quently, all energy dissipation was observed in the connections between thetimber panel and rigid steel beam.Soon after the one wall panel tests, a full scale three storey CLT buildingwas built and tested at the National Institute for Earth Science and DisasterPrevention (NIED) Tsukabu shake table facility in Japan (Ceccotti andFollesa, 2006; Ceccotti, 2008). Test results further confirmed the reality ofthe stiff CLT panels. When an 0.8g scaled Kobe ground motion was appliedto the building small deformations were found in the screws between thepanels at the vertical point. These deformations in the screws continued with233.2. CLT buildingssequential increasing earthquake tests until a hold-down failure occurred andfurther deformation took place in the vertical joint screws between panels.Using the experimental tests from the single wall panel, the connections werecalibrated in a finite element model and then modeled to the specificationsof the three-storey CLT building tested. Comparisons of the model andexperiment showed good agreement (Ceccotti, 2008).Following the three storey building study, a seven storey CLT buildingwas tested at NIED’s shake table in Miki, Japan (Ceccotti, 2010; Ceccottiet al., 2013). Details of the seven storey building can be seen in Figure 3.3with minor changes between floors maintaining a consistent load path.(b) (a) Figure 3.3: 7 storey CLT experimental test: a) building set up; b) plans forbuilding ([Ceccotti et al., 2013] by permission from publisher)The shake table tests validated the importance of connections within tim-ber buildings. Furthermore, results showed that the CLT system had a veryhigh stiffness; however, the system also had good ductility and dissipatingmechanisms (Ceccotti et al., 2013). Failure modes observed were ductile,occurring locally to the fastener in bending and embedment as shown inFigure 3.4. Moreover, after all 10 major ground motion tests, the buildingshowed no residual displacement (Ceccotti et al., 2013). However, the low243.2. CLT buildingsresidual drift was a result of the extremely high stiffness of the CLT build-ing. This stiffness resulted in accelerations of 4g at the top storey duringthe intense ground motions (Quenneville and Morris, 2007). Although thebuilding survived with relatively no damage, the high acceleration demandwill be detrimental for non-structural failures.(b) (a) (c) Figure 3.4: 7 storey CLT building connection failures: a) hold-down fastenerfailure; b) nail pull-out in bracket; c) compression perpendicular to grain inwood failure ([Ceccotti et al., 2013] by permission from publisher)One of the main problems that reoccurred in the SOFIE projects was thehold down connections to the concrete foundation. A team of researchers inItaly carried out experimental tests on the cyclic behaviour of these hold-downs through experimental tests (Gavric, Fragiacomo, and Ceccotti, 2014).Gavric et al. (2014) found the yield strength of the strongest hold-downbracket to be 40.46 kN. Furthermore, Polastri, Pozza, Trutalli, Scotta, andSmith (2014) discussed that pure CLT buildings over five stories did notseem possible with current hold-down capabilities for high seismic locations.3.2.2 FPInnovations CLT testingA team of researchers at FPInnovations developed the CLT handbook,one of the most rigorous works on CLT to date (Gagnon and Pirvu, 2011).The handbook provides users with a guide to the manufacturing, struc-tural design, seismic performance, connection design, duration of load and253.2. CLT buildingscreep effect, vibration performance, fire performance, acoustic performance,building enclosure design, and lastly the environmental performance of CLT.Most important to this research is the chapters published on structural de-sign, seismic behaviour and connection design (Gagnon and Pirvu, 2011).The seismic behaviour chapter includes experimental studies on the lateralperformance of CLT walls subject to both monotonic and cyclic loads asshown in Figure 3.5.(b) (a) Figure 3.5: FPInnovations CLT wall test: a) test apparatus; b) panel duringtesting ( c© 2011, FPInnovations, by permission)Results were similar to that seen in the SOFIE projects where failurewas seen in the connections as the CLT panel was much stiffer in relation.Figure 3.6 depicts the bracket connections between the steel beam and CLTpanel with parallel-to-the-grain (Figure 3.6a) and perpendicular-to-the-grain(Figure 3.6b) loading. The hold-down connection and angled bracket walltest is shown in Figure 3.7.The bracket connections were further studied under reverse cyclic loadingat FPInnovations and the results are discussed in various papers (Schneideret al., 2014; Shen, Schneider, Tesfamariam, Stiemer, and Mu, 2013). Thesetests provided a significant amount of knowledge of the bracket connectionsused to connect the steel beam to the CLT. Results showed that nail pull-out was the dominant failure mode; however, CLT crushing was also seen insome instances. Analytical models were developed for the bracket connec-263.2. CLT buildings(b) (a) Figure 3.6: FPInnovations connection test: a) parallel-to-the-grain; b)perpendicular-to-the-grain ( c© 2015, Johannes Schneider, by permission)Figure 3.7: FPInnovations CLT wall test showing angled connection (left)and hold-down connection (right) ( c© 2011, FPInnovations, by permission)273.2. CLT buildingstions using OpenSees and showed good correlation when compared with theexperimental work (Shen et al., 2013).3.2.3 University of Canterbury UFP post-tensioned timbercore wallsResearch at the University of Auckland on PREcast Seismic StructuralSystem (PRESSS) concrete walls influenced researchers at the Universityof Canterbury to apply the knowledge from PRESSS to mass timber shearwalls (Dunbar, 2014). CLT panels were utilized as the shear walls in thecore system and timber perimeter beams surrounded these panels. Low andhigh seismic design parameters were both considered and subsequently, twoconfigurations were designed. The perimeter beams were fastened to thecore using screwed and epoxied steel plates (high seismic specimen) or aring of bolts (low seismic option). Furthermore, the CLT core walls werepost tensioned using 7-wire strands and tested with different magnitudes oftensioning. Connections in the high seismic specimen were made betweenthe CLT panels and steel columns using U-shaped flexural plates (UFP’s).The UFP plates dissipated energy and provided constant maximum strainvalues under positive and negative displacements. Figure 3.8 displays theCLT post tensioned core walls with timber perimeter beams and UFP platedesign method proposed by Dunbar (2014).Full scale experimental tests were done on the low and high seismicoption timber core wall systems. Results of the study proved that the post-tensioned coupled walls provided a significant increase in hysteretic dampingover the single walls as shown in Figure 3.9. However, most of the energydissipation in the system was seen to be between the adjacent wall pan-els. Under high seismic load significant friction force was developed in thecoupled walls resulting in a very stiff and strong system.These results also showed that the UFP devices had a low influence onthe hysteretic behaviour of the CLT coupled walls. Minimal damage wasobserved in the experimental tests but crushing at toes of coupled walls anddeformation in the replaceable UFP device was noted. This experimental283.3. Hybrid buildings(b) (a) Figure 3.8: CLT core walls with post tensioning and UFP connections: a)3D rendering of high seismic core option; b) UFP plate ( c© 2014, AndrewDunbar, by permission)study showed good seismic behaviour of the CLT core system with UFPdevices displaying effective energy dissipation and re-centering capabilities.3.3 Hybrid buildingsPast research has demonstrated the benefits of hybridization and tim-ber core systems. The most relevant studies are presented in the followingsections.3.3.1 SOM tall timber projectSkidmore, Owings and Merrill (SOM) Structural Engineers released areport on the conceptual design of tall (up to 42 storeys) hybrid timber-concrete buildings (SOM, 2013a). The benchmark building was a 42 storeyconcrete core building in Chicago designed by the company in 1966. Theproposed design Concrete Jointed Timber Frame utilized a timber core sys-293.3. Hybrid buildingsFigure 3.9: Coupled and single wall comparison ([Dunbar et al., 2014] bypermission from publisher)tem and concrete spandrel beams. Environmental impacts due to concreteand steels carbon emissions were the main reason for the research into the re-port as the report states structural systems that minimize embodied carbonfor tall buildings allow the positive environmental aspects of tall buildingsto be more pronounced (SOM, 2013a). Therefore, the purpose of the reportwas to develop a system that uses mass timber as the main structural ma-terial capable of resisting forces up to a 42 storey height. A typical floorstructure of the Concrete Jointed Timber Frame can be seen in Figure 3.10.Concrete link beams provided the necessary ductility under lateral loadstherefore allowing the timber to remain in an elastic state. The reportshowed that as concrete beams have much greater shear strength when com-pared with timber beams, it is a more efficient choice. Furthermore, theyshow that 24x 24 (610 x 610 mm) timber columns are capable of carrying the1,200 kips (5338 kN) of axial compression typically required in tall buildings.The design engineers from SOM contend that the proposed ConcreteJointed Timber Frame will behave as predicted under extreme lateral load.303.3. Hybrid buildingsREINFORCED CONCRETE LINK BEAMSTIMBER FRAMING WITHIN CORETYPICAL FLOOR STRUCTUREBUILT UP TIMBER COLUMNSSOLID TIMBER SHEAR WALLSSOLID 8” THICK TIMBER FLOOR PANELSREINFORCED CONCRETE SPANDREL BEAMREINFORCED CONCRETE WALL JOINTFigure 3.10: SOM’s concrete jointed timber frame typical floor structure ( c©2013, Skidmore, Owings & Merrill LLP, by permission)However the engineers noted several structural, architectural, fire perfor-mance, construction and manufacturing concerns that must be both re-searched and physically tested before the proposed building could be built.3.3.2 mgb architecture + design and equilibrium consultingArchitect Michael Green collaborated with engineer Eric Karsh fromEquilibrium Consulting of Vancouver, B.C. to produce a report on tall tim-ber buildings, released in 2012: The Case for Tall Wood Buildings(Greenand Karsh, 2012). The comprehensive report includes detailed discussionon the need for tall timber buildings as well as a case study design. The col-laboration suggested a new design method using mass timber panels termedFFTT (finding the forest through the trees). Mass timber panels such asCLT, laminated strand lumber and laminated veneer lumber were studiedin the report. Defining the FFTT system was a timber-steel hybridization313.3. Hybrid buildingsutilizing mass timber panels for the vertical structure, lateral shear wallsand floor slabs. Additionally, steel beams connected the mass timber panelsand provided the necessary ductility. Four options were proposed for theFFTT system by the researchers Green and Karsh with the most rigorousoption reaching 30 storeys in height for the timber core, interior shear wallsand exterior moment frame. Figure 3.11a shows the plan view of option 1 ofthe FFTT system with mass timber core walls. Furthermore, Figure 3.11bshows an exploded view of the core system proposed in the FFTT system.In comparison with the benchmark concrete building, cost analysis of theFFTT system was done. Results of the cost analysis showed the FFTTsystem not only as feasible but as cost competitive when compared to thebenchmark concrete building.(b) (a) Figure 3.11: FFTT system: a) Plan view; b) 3D rendering of core explodedview ([Green and Karsh, 2012] by permission through Creative CommonsLicense)Researchers at UBC (Bhat, 2013; Fairhurst, 2014; Zhang et al., 2015)studied the structural behaviour of the proposed FFTT system. Bhat (2013)conducted experimental tests on the timber core wall to steel beam connec-tion as shown in Figure 3.12. Moreover, Fairhurst (2014) analyzed the pro-posed four FFTT options using the finite element program OpenSees with323.3. Hybrid buildingsthe calibrated connections from Bhat’s study (2013).(b) (a) Figure 3.12: FFTT system: a) Connection detail ([Green and Karsh, 2012]by permission through Creative Commons License); b) Full embedment steelbeam to CLT wall experimental test ( c© 2013, Pooja Bhat, by permission)The research showed good seismic behaviour of the FFTT system understatic and dynamic loading (Zhang et al., 2015). Results indicated that thetaller structures governing load case was due to wind loads rather than seis-mic loads because of the reduced mass of the structure from the lightweighttimber products (Fairhurst, 2014). However, the FFTT study displayed thattimber-steel hybrid structures are capable of resisting large lateral loads fromearthquakes with a ductility factor (Rd) of 5 while limiting the inter-storeydrift to 2.5% (Zhang et al., 2015).3.3.3 UBC CLT infilled SMRFRecently at the University of British Columbia (UBC) researchers havestudied a new building method utilizing CLT walls as infill within steelmoment resisting frames (Bezabeh, 2014; Dickof et al., 2014; Dickof, 2013;Schneider, 2015; Tesfamariam et al., 2015; Tesfamariam et al., 2014). TheCLT infill walls were intended to provide hysteretic damping during se-vere ground motions through crushing behaviour. L-shaped brackets testedat FPInnovations as shown in Figure 3.6 connected the steel beams and333.3. Hybrid buildingscolumns to the CLT panel. A gap was detailed between the CLT panel andthe steel beams and column in order to allow for deformation within thebrackets and then crushing of the timber panels. The gap can be seen inFigure 3.13 as can the proposed design method and elevation view of thesix-storey, three-bay building with two bays infilled with CLT.Figure 3.13: CLT infilled steel moment resisting frame: a) proposed method;b) elevation view of two bays infilled six-storey structure ( c© 2013, CarlaDickof, by permission)Results of the studies showed the benefits of this hybridization of steeland timber withstanding extreme ground motions. Detailed finite elementanalysis was done on the timber-steel hybrid system using monotonic pushoveranalysis to determine the overstrength and ductility factors (Dickof et al.,2014; Tesfamariam et al., 2015). The researchers recommended that a duc-tility factor (Rd) of 4 and overstrength factor (Ro) of 1.5 would yield asafe and economical design for their proposed structure (Tesfamariam et al.,2015). Furthermore, nonlinear time history analysis (NLTHA) and incre-mental dynamic analysis (IDA) were used to validate the proposed seismicmodification factors. The hybrid structure has shown to be effective in com-bining the ductile and strong steel moment frames with lighter and stiffCLT infill walls. Additionally, the provided L-shaped brackets ensure fullconfinement between the structural elements and energy dissipation under343.4. Representative buildingsintense seismic shaking. Furthermore, Bezebah (2014) has shown that theenergy dissipation of the CLT panel is effective in reducing interstorey driftto within 2.5%. This reduced drift ensures building safety following a seismicevent and allows for quick and simple repairs (Bezabeh, 2014).3.4 Representative buildingsSeveral significant mass timber projects are currently under constructionin North America that will set new significant achievements and are worthnoting. The recently finished Wood Innovation and Design Centre in PrinceGeorge, B.C. is currently the world’s tallest modern all-timber structure(Michael Green Architecture, 2016b). Standing 6 storeys high the structureis made up of CLT, laminated veneer lumber, and glulam beams and columns3.14) (Michael Green Architecture, 2016b).Figure 3.14: Wood Innovation and Design Centre ( c© 2016, Ema Peter, bypermission)Michael Green Architecture and the DLR group designed a 7 storey office353.4. Representative buildingsbuilding in Minneapolis (T3 Minneapolis) consisting of one floor of concreteand 6 pure timber floors above the concrete(Figure 3.15). T3 Minneapolisutilizes a concrete core, nail laminated timber floors and a glulam post andbeam frame. The building is scheduled to be constructed by the fall of 2016.Figure 3.15: T3 Minneapolis ( c© 2016, Michael Green Architecture, by per-mission)In Vancouver, B.C. an 18 storey residence tower (Brock Commons) isbeing constructed at the University of British Columbia. Similarly to T3Minneapolis, the lateral system is a concrete core wall. CLT floors andglulam columns control the gravity loads and transfer the shear loads backto the core (Figure 3.16). Once completed the building will be the one ofthe tallest mass timber buildings in the world. Brock Commons is set to becompleted in the fall of 2017.363.4. Representative buildingsFigure 3.16: Brock Commons rendering ( c© 2016, Acton Ostry Architects,by permission)37Chapter 4Steel-timber core wallsThe timber-core wall system was developed to display that CLT panelscan be designed to resist lateral loads when hybridized with steel compo-nents. As discussed in Chapter 2, past CLT studies have shown that CLTbuildings with currently available steel bracket and hold-down connectionsare inadequate in high seismic regions. CLT is a great structural productbut without hybridization the product can’t reach its full potential in seis-mic applications. The aim of this thesis is to develop a system that utilizedthe stiffness of the CLT to resist the seismic forces from a severe earth-quake in a structural building core system. A hybrid system was developedthat couples the light and stiff CLT walls with steel plates and ductile steelconnections.4.1 Lateral resisting systemThe steel plates were designed to run the height of the wall and fastenedtogether by screws. These walls were connected at the intersection of eachfloor by t-stub connections that are bolted together as shown in Figure4.1. The core walls land on the floors below therefore allowing for quickconstruction as the platform construction method allows the builders tohave a safe floor to work on while installing the next one, saving money andtime.As the figure shows, the steel plates were not connected to the floors asthe floors connect to the core walls with brackets and hold-down connections.The seismic forces were transferred to the core wall through the diaphragmand thus the connections of the diaphragm to core walls. Furthermore, it isimportant to note that the designed platform construction style resulted in384.1. Lateral resisting systemFigure 4.1: 3D rendering of bolted t-stub end plate connectionthe walls landing on top of the steel-timber core walls. The proposed hybridsystem was developed after several iterations of core systems and was foundto be the most capable CLT-steel hybrid core wall system in high seismiczones. The following sections detail the connections of the timber-steel corewall hybrid system.4.1.1 T-stub connectionThe t-stub connection provided the plastic behaviour of the timber corewalls. Under seismic action all other components of the building design weredesigned to remain elastic while the t-stub is designed to behave plastically.Therefore, the t-stub connection dissipated the seismic force and providedductility to the stiff system. Researchers Piluso, Faella and Rizzano (2001)studied this T-stub connection and validated their theories with experimen-tal work. Figure 4.2 shows the structural details of the t-stub connectionsthat were tested by the researchers. Connecting the two end steel FE430plates were high strength 27 mm diameter bolts.Figure 4.3 displays one of the t-stub connection results from the tests.394.1. Lateral resisting systemFigure 4.2: Structural details of the tested specimen ([Piluso and Rizzano,2008] by permission from publisher)The test subject developed four plastic hinges including: the two boltsand two in the steel end plates at the yield formation lines. The force-displacement test with increasing load on the connection displays the ben-efits of this connection system. The connection is very stiff until its yieldpoint. After yielding, the connection plastically deforms at a consistentslope until complete failure.Cyclic tests were conducted in Italy by researchers Piluso and Rizzanoon the T-stub connection (Piluso and Rizzano, 2008). The cyclic tests werecarried out under displacement control with a monontonic test, constantamplitude cyclic tests and variable amplitude cyclic tests. The study showedgreat cyclic behaviour of the specimens as shown by the variable amplitudecyclic test in Figure 4.4. The test shows no negative deformation as thet-stub connection is extremely stiff in compression due to the steel platecompressing into the other steel plate. Negative deformation would be aresult of the steel plate buckling above or below the connection. However,this timber-core system was designed to eliminate the possibility of bucklingin the steel plates.404.1. Lateral resisting systemFigure 4.3: Experimental tension testing results for t-stub connection test 6([Piluso and Rizzano, 2008] by permission from publisher)(b) (a) Figure 4.4: Experimental cyclic tests on t-stub connection: a) yielding inwelded connection; b) force-displacement curve results for test B5 ([Pilusoand Rizzano, 2008] by permission from publisher)4.1.2 BracketsBoth L-shaped brackets and conventional CLT hold-down brackets wereutilized in the hybrid core system to transfer the axial and shear forces fromthe diaphragm to the core. The shear forces were a result of the seismicforce and the axial force was a result of the shear and gravity load transfer414.2. Behaviour of timber-steel core wallsrequired of the core walls. Details of these connections can be found inChapter 3 (Gavric et al. ,2014; Polastri et al., 2014; Schneider et al., 2014).It is important to note that the connections were designed to remain elasticunder both gravity and lateral loading events.4.2 Behaviour of timber-steel core wallsLateral behaviour of the timber-steel core walls was defined throughfirst principles and verified using finite element software. A one storey twodimensional shear wall of the core system was defined as shown in Figure4.5. The shear and bending deformation (∆2) is outlined by the dotted linesand the anchorage deformation (∆1) is outlined by the solid lines. Moreover,the deformation of the connection at B is defined by ∆T . The dimensionsof the core wall system are defined as the height (H), length (L), effectivelength (Le), and the distance between the edge of the shear wall and theconnection is defined as d3. Loading is defined by P acting at the top of thewall representing the force from the diaphragm above due to the earthquake.424.2. Behaviour of timber-steel core walls∆1 ∆2 𝐻 𝐿𝑒 𝑑3 𝐿 ∆𝑇 𝐴 𝐵 𝑃 Figure 4.5: Core system deformationA free body diagram for the core system is shown in Figure 4.6. Theloading force on the shear panel is due to the brackets connecting the CLTfloor to the CLT shear panel below. The forces at the base (FA and FB)represent the forces seen by the t-stub connections at points A and B.434.2. Behaviour of timber-steel core walls13P FA FB 13P 13P Figure 4.6: Free body diagram of timber-steel core systemThrough system equilibrium the reactions at A and B were defined basedon the input force (P ) and the system dimensions:P =FB(L− d3)H(4.1)Moreover, in a similar manner the elastic deformation of the system dueto connection deformation (∆T ) was computed:444.2. Behaviour of timber-steel core walls∆T =PHk(L− d3)(4.2)where k is the stiffness of the t-stub end plate hold-down connection.Similarly the deformation of the connection was related to the storey defor-mation:∆1 =∆THL− d3(4.3)Substituting Equation 4.2 into Equation 4.3 for the connection deforma-tion resulted in the connection storey deformation in terms of input force,height, length and stiffness:∆1 =PH2k(L− d3)2(4.4)The total deformation (∆) of the one storey shear wall system was dueto the bending, shear and connection slip deformation. The deformationswithin the panel were due to flexural bending (∆b) and shear bending (∆s):∆2 = ∆b + ∆s (4.5)where the bending and shear deformation can be quantified using theequation below for shear walls:∆2 =PH33EI+1.2PHGA(4.6)where E is the modulus of elasticity of the wall, I is the moment ofinertia of the wall, 1.2 is the shape factor used for shear walls and G is the454.2. Behaviour of timber-steel core wallsmodulus of rigidity of the wall.The deformation within the wall was based off a typical shear wall asneither steel or timber shear walls were an adequate fit for the system. As thepanel is an orthotropic shell the CLT shear wall and steel plate hybridizationact together. However, because the shear wall was not fixed at the corners ofthe shear wall a factor was proposed (a) to account for the reduced restraintat the base:a =LLe(4.7)Figure 4.5 shows the total deformation as the addition of the bendingand shear deformation with the connection deformation:∆ = ∆1 + ∆2 (4.8)Using Equation 4.4 and Equation 4.6 the total deformation was deter-mined:∆ =aPH33(ECLT ICLT + EstlIstl)+1.2aPHGCLTACLT+PH2k(L− d3)2(4.9)When the yield point of the connection was reached, the stiffness (k)began to drop and the connection began deforming plastically. Past the yieldpoint of the connection, the CLT remained behaving elastically assumingthe stresses remain below the yield point of the CLT panel. For this reason,during post processing the CLT panel maximum stress was checked to ensurethe panel remained elastic before failure of the steel hold-down connection.The equation was validated using finite element software. The steel framemembers were connected to the CLT shear wall at every 100mm to ensurethey behaved as a unit. However, the results saw negligible differences whenthe spacing was increased to 300mm.464.2. Behaviour of timber-steel core wallsThe steel-timber core wall system was designed to be implemented inmulti-storey buildings. Therefore, the lateral behaviour of the core wallswas extended into a multi-storey one bay two dimensional systems as shownin Figure 4.7.Δ Figure 4.7: Multi-storey timber-steel core-wall systemTo determine the deformation of the multi-storey timber-steel core wallsystem the equations previously defined were used in combination to accountfor the increasing moments due to additional loads from storeys above. The474.2. Behaviour of timber-steel core wallsshear and bending terms from Equation 4.9 remain relatively unchanged;the anchorage deformation however was adapted to account for the increas-ing load from above storeys. Equation 4.10 was developed through firstprinciples to find the deformation due to the deformation of the t-stub con-nections.∆1 =[(∑Ni=1 Pi)Hi +(∑Ni=1 Pi+1)Hi+1 +(∑Ni=1 Pi+2)Hi+2]Hik(L− 2d3)2(4.10)The deformation from the t-stub connections allows the timber-steelpanel to rotate unlike the shear and bending in the panel that only resultsin minor negligible rotation. The deformation from the t-stub connectionswas used to define this resulting angle of rotation. The rotation angle iscumulative as you go up the height of the building. Therefore, the final de-formation of a particular storey in the multi-storey timber-core wall systemcan be found with the deformation from the shear, bending and anchorageand the angle of rotation. Equation 4.11 defines how to find the deformationin the relative storey; Equation 4.12 defines how to find the rotation anglefor the storey (θN ) based on the rotation angle from the previous storey(θN−1); and Equation 4.13 defines how to find the rotation for the relativestorey (θi).∆N = H · sinθN + (∆1 + ∆2) (4.11)θN = θN−1 + θi (4.12)θi = tan(∆1Hi)(4.13)484.2. Behaviour of timber-steel core wallsWith this knowledge, the total deformation was determined for eachstorey. The equation was taken one step further and developed to account forthe nonlinear behaviour of the t-stub connection. As the nonlinear behaviourof the system is restricted only to the t-stub connections, using the sameequations the deformation can be defined nonlinearly by adjusting just thedeformation due to anchorage. This nonlinear method was developed andused for the design.The first principles method was analysed by applying the same loadbased off the ESFP on the structure using lateral point loads on the devel-oped method and SAP2000 model. SAP2000 is a finite element programand is discussed in Chapter 5. The displacement at each storey is displayedin Figure 4.8 for the developed first principles method and SAP2000 model.Results show that the method is accurate and will predict the behaviour ofthe timber-steel core wall structure well. Plastic deformation was observedin the first three floors t-stub connections with first principle analysis. Theremaining upper floor t-stub connections remained elastic under the lateralloading.494.2. Behaviour of timber-steel core walls0123456780 100 200 300 400 500Storey Deformation (mm) First principles SAP2000Figure 4.8: Storey drift of first principles and SAP2000 timber-steel corewall system50Chapter 5Multi-storey building designThe timber-steel core wall system was utilized in the design of a multi-storey 7-storey benchmark building. The benchmark building was a 7-storeyresidential concrete building located in Vancouver, Canada. The buildingfloor plan is shown in Figure 5.1. The building has structural irregularitiesallowing for comparisons between the built benchmark building and pro-posed timber-steel hybrid building. The floor plan remained constant fromfloors 1 through to 7; however, a mechanical room on the roof was includedin the design similar to the benchmark building. The height of the firststorey was 3.6m with each floor above having a height of 3m resulting in thetotal building height (including the mechanical roof top room) of 24.6m.BEAMSCORE WALLSCOLUMNSCLTPLATES25.14 m16.62 mSTAIRWELLSTAIRWELLELEVATORFigure 5.1: Design building floor plan515.1. Gravity loadingAs the floor plan shown in Figure 1, there are three main core elementsin the building including two stairwells and one elevator core. Several di-mensional irregularities in the benchmark building were accounted for in thisstudy. Where doors and windows were specified, openings were introducedto ensure an accurate comparison with the concrete benchmark building.Therefore, one main goal during the design process was to choose practi-cal, standard and constructible connection details. Relevant design loadsand climatic information for the building is shown below in Table 5.1. Theductility factor Rd = 2.0 and overstrength factor Ro = 1.5 were based onrecommendations CLT Handbook (Gagnon and Pirvu, 2011).Table 5.1: Design loads and climatic information for the benchmark buildingLoads Roof Ss = 1.8 kPa and Sr = 0.2 kPa, Dead Load = 2.0 kPa 3rd - 9th Floor Live Load = 1.9 kPa, Dead Load = 2.0 kPa Storage and Corridors Live Load = 4.8 kPa, Dead Load = 0.3 kPa Seismic Climatic Data Sa(0.2) = 0.94, Sa(0.5) = 0.64, Sa(1.0) = 0.33, Sa(2.0) = 0.17, PGA = 0.46, Site Class 'C' Importance Factor IE = 1.0 Force Modification Factors Rd = 2.0, Ro = 1.5 5.1 Gravity loadingAlthough the behaviour of the gravity system were not the focus of thisthesis, the system was sized following the requirements of the NBCC 2010in order to find a reasonable weight for the analytical models (NRC, 2010).The steel-timber core walls were designed to transfer gravity load as well asthe lateral load in the building; however, as the smallest size of panels wouldhave been adequate in resisting the gravity load the design of the system iscontrolled by the lateral design. For this reason the steel-timber core systemgravity design was not studied until later in the design process.525.1. Gravity loading5.1.1 Floor systemThe floor panels were designed to stretch across a maximum span of6 metres and behave similarly to a one-way slab. Moreover, these CLTfloor panels included a 1 inch concrete topping in order to satisfy vibrationcriteria. Strength criteria of the CLT panel for the largest gravity loadcombination was met using 5 layer CLT floor panels.5.1.2 DiaphragmDue to the buildings irregularity, a rigid diaphragm was necessary totransfer the loads to the core walls. Recent studies (e.g. Ashtari, Haukaas,and Lam, 2014; Fast, 2014; Moore, 2000) on in-plane forces acting in CLTpanels have given the much needed insight into the diaphragm behaviour.Moore (2000) explained the design of a hybrid timber-steel flooring sys-tem for a 12 storey apartment building constructed in 1999. Soon after,researchers began exploring the behaviour of CLT panels loaded in plane.Ashteri et al. (2014) found that the rigidity of the diaphragm highly de-pends on the stiffness of the buildings shear walls. Moreover, researchershave pointed out that assuming a rigid diaphragm for a CLT floor is nottotally accurate (Fragiacomo, Dujic, and Sustersic, 2011). However, the re-searchers do note that rigid diaphragms are possible by gluing adjacent floorpanels together or by using a very stiff connection between the two panelssuch as screws at 45 degree angles.5.1.3 Beams and columnsThe steel columns and beams were sized according to the maximumaxial force when the structure was loaded following the NBCC 2010 loadcombinations. Member sizing can be seen below in Table 5.2.535.1. Gravity loadingTable 5.2: Steel gravity frame design detailsFloor Section Columns All HSS 152x152x4.8 Beams All W 310x24 In a detailed design process for construction the size of columns wouldbe optimized for the floor. However, as the columns and beams were notessential parts of the study the members remained constant through allthe floors to simplify the design. The steel columns were assigned momentreleases at either end as they are only responsible for the gravity load system.In contrast, the steel beams were continuous and fully attached to the CLTfloor panels.5.1.4 ConnectionsSeveral connections were necessary in designing the gravity system con-necting timber to timber, steel to steel and hybrid connections connectingsteel to timber. All connecters were made of steel and designed to be ductileor stiff depending on the desired effect. Options for each connection werestudied and presented in the following sections.Panel to panelConnecting panels along their longitudinal edges can be done in a va-riety of ways as shown in Figure 5.2. There are more options available tothe designer; however, these two connections are most common. Moreover,these connections can be modified in a number of ways to produce the mostefficient connection for a particular case. For example, by adding a layerof plywood that overlaps both panels and securing it to either panel theductility of the system can be increased (Fragiacomo et al., 2011). However,the plywood strip reduces the stiffness of the screwed connection. The dou-ble surface spline is a very stiff and strong connection between wall or floor545.1. Gravity loadingpanels. Unfortunately, as the detail shows, the work required to achieve theconnection in construction is costly. Contrary to the double surface splineconnection, the half-lapped joint with self-tapping screw connection allowsfor quick installation and can provide good strength and stiffness values.(b) (a) Figure 5.2: CLT to CLT wall or floor connection: a) double surface spline;b) half-lapped joint ( c© 2011, FPInnovations, by permission)Fragiacomo et al. (2011) achieve the necessary strength for their CLT toCLT panel connection using a 6mm diameter self-tapping screws at a 300mmspacing centre-to-centre in a 50 mm step joint. Their studies utilized theconnections test data reported in Gavric et al. (2012). Experimental resultsof the tested lap joint connection can be seen below in Figures 5.3 and 5.4.555.1. Gravity loading(b) (a) Figure 5.3: CLT floor panel connection test setup ([Gavric et al., 2012] bypermission from conference publishers)(b) (a) Figure 5.4: Hysteresis loops of lap joint test ([Gavric et al., 2012] by per-mission from conference publishers)565.1. Gravity loadingBased on the discussed research and testing done by MiTiCon Connec-tors the connection between CLT floor panels was designed to be stiffer thanthe CLT panels. By designing the connection this way the diaphragm canbe modeled as a continuous floor system without connectors. This ensuresthe model design is conservative by taking the smallest in-plane stiffness ofeither the CLT panel or connection. To achieve such a stiff connection ahalf-lapped joint was utilized with self-tapping high strength screws. Theconnection required four SWG ASSY VG screws per meter at 45 degreesin each direction at 50 mm on centre spacing between screws of oppositeorientation.Panel to beamSteel beams were designed to transfer the gravity loading from the tim-ber floor panels to the steel columns. Connecting two different buildingmaterials is challenging in any system and requires more detail than typicalconnections. Fortunately, much research lately has focused on timber-steelconnections (e.g. Asiz and Smith, 2011; Loss et al. 2014; Schneider etal. 2014; Shen et al., 2013). Experiments were conducted in Italy (Losset al., 2014; Loss, Piazza, and Zandonini, 2015a, 2015b) to determine thebehaviour of CLT panels secured to wide flange steel beams. The study con-sidered several different connections between the beam and panel and theresults were presented and discussed. Most notably test # five in their studyconsidered inclined screws into the CLT panels from angles that were weldedto the steel beam. The results showed ductile behaviour of the connectionsbetween the steel beam and CLT panel after yielding as shown in Figure 5.5.The connection between the steel beam and floor panels, however, was de-signed to remain elastic; therefore, the connection was not studied in detailfor this research.575.1. Gravity loading(b) (a) Figure 5.5: CLT panel to steel beam connection: a) testing set up; b)results of monotonic and cyclic test ([Loss et al., 2015a] by permission frompublisher)Beam to columnA simple load bearing connection was proposed for the connection be-tween the steel beam and steel column. The connection was designed tostay well within the elastic range and transfer the axial and shear forcesfrom the beam to the columns. Numerical values for the connection weretaken from the Canadian Institute of Steel Construction (CISC) Handbook(CISC, 2010). Several other connection options are available to transfer thegravity load from the wide flange beams to square hollow columns and canbe found in the AISC Design Guide 24 for hollow steel sections (Packer,Sherman, and Lecce, 2010).Beam to wallThere are a few instances in the designed building where a steel beamconnects with a CLT core wall panel. To transfer the loads between the twoelements an end plate was proposed. The end plate would be welded to thesteel beam and use high strength screws placed at an angle using the angledwedge washers (Figure 5.6). This connection would transfer the necessary585.2. Modeling of lateral and gravity loading resisting systemsshear forces to the CLT shear wall.(b) (a) Figure 5.6: High strength screws: a) 8, 10 and 12 mm VG CSK ASSYScrew types with associated angled wedge washer; b) ASSY screw connectingsteel plate to timber member ( c© 2014, MiTiCon Timber Connectors, bypermission)5.2 Modeling of lateral and gravity loadingresisting systemsTo validate the seismic design of the building finite element software isused by structural engineers to predict the response. The analysis in thisthesis was carried out using the SAP2000 commercial finite element softwaredeveloped by Computers and Structures Inc. (CSI, 2013). The programapplies the desired loading on the input parameters loaded by the designer.The main advantage of finite element software in structural engineering is theability of the software to account for the nonlinear capacity of the materials.Furthermore, the program allows users to apply dynamic loading such asearthquake ground motions on the structure to analyze the response underextreme loading. The following sections outline modelling details of thesteel-timber hybrid building Figure 5.7.595.2. Modeling of lateral and gravity loading resisting systemsCLT CORE-WALL STEEL PLATE T-STUB BRACKET CLT FLOOR V Figure 5.7: Model of proposed timber-steel core wall system5.2.1 CLT panelsCLT panels are manufactured using odd numbered layers; currently 3,5, 7, and 9 layer CLT panels are available for production by the most pop-ular CLT manufacturers (KLH, 2012; Structurlam, 2013). For this reason,even though the strength in each layer is the same, with an odd number oflayers there is always one more layer in one direction making it the strongerdirection. Figure 5.8 shows the different axis of strength of an orthotropicCLT panel. Axis 2 is the strong axis in bending in this case as it has two ofthe three layers running lengthwise. The first axis is weaker than the secondaxis in bending and the third axis is the weakest as the depth of the memberin this direction is small compared with axes 1 and 2. The CLT panels weremodeled accordingly as orthotropic shell elements. To determine the modu-lus of elastic and shear moduli of the CLT panel in all planes the ’k Method’was used from the CLT Handbook (Blass and Fellmoser, 2004). The k fac-605.2. Modeling of lateral and gravity loading resisting systemstors depend on the Young’s modulus or shear modulus in the considereddirection and at ninety degrees to the direction considered. The orthotropicCLT panel k factors required were calculated using the formulas as shownin Table 5.3.1 2 3 Figure 5.8: Orthogonal axes of a CLT panel615.2. Modeling of lateral and gravity loading resisting systemsTable 5.3: K factors for CLT ( c© 2011, FPInnovations, by permission)625.2. Modeling of lateral and gravity loading resisting systemsFollowing the Structurlam design guide the Young’s modulus and shearmodulus were developed in each direction according to the composite theory,’k Method’. The initial values for Young’s modulus and shear modulusfrom the manufacturer were multiplied by the k factor which equated to theproperties in their respective direction.SAP2000 uses three node triangular elements or four node quadrilateralelements to solve finite element shell problems (CSI, 2013). However, thefour node quadrilateral elements are more accurate and will be used in thisstudy as all elements are rectangular. The shell’s stiffness is calculated using2x2 Gauss integration points and then extrapolated on to the element joints.The four node quadrilateral elements nodes are numbered according to thelocal coordinates as shown in Figure 5.9.Figure 5.9: Four node quadrilateral shell element ( c© 2013, Computers andStructures, by permission)In this building 5 and 7 layer CLT panels were designed. Following the ’kMethod’ the orthotropic properties were developed and are summarized inTable 5.4. The k factors were computed using the elastic and shear modulusin both parallel (E0, G0) and perpendicular (E90, G90) directions and thedimensions of the layers (α) as explained in the bottom of the table.635.2. Modeling of lateral and gravity loading resisting systemsTable 5.4: Orthotropic CLT propertiesYoung's Modulus Shear Modulus Eo = 9500 MPa Go = 950 MPa E90 = 317 MPa G90 = 50 MPa 5 Layer E1 = 3794 MPa G1 = 379 MPa E2 = 6022 MPa G2 = 602 MPa E3 = 258 MPa G3 = 41 MPa 7 Layer E1 = 4275 MPa G1 = 428 MPa E2 = 5539 MPa G2 = 554 MPa E3 = 231 MPa G3 = 36 MPa 5.2.2 Steel membersAll steel members were input to the model using the SAP2000 databasefor Canadian steel members. However, the plates were input to the modelthrough the section designer provided in SAP2000. The frame members werediscretized according to the accuracy necessary for the study. However, themeshing of the steel members has shown to have a small effect on the results,so a larger mesh size was appropriate to save computing effort.5.2.3 ConnectionsAccurately modeling the connections is the most critical detail in theanalysis of the steel-timber hybrid building and dictates whether the ob-tained results were correct. With extensive research into the program anal-ysis methods the connections were modeled appropriately. Using experimen-tal tests the connections were validated using SAP2000 and then comparedwith experimental work. Elastic and nonlinear spring elements were utilizedin the building model.Two node spring elements were used within the analytical model to de-pict the behaviour of the steel connections between the CLT shear wall and645.2. Modeling of lateral and gravity loading resisting systemssteel plate and CLT wall to floor connection. Linear and multilinear plas-tic springs were used depending on the connection strength and role in thebuilding. Nonlinear plastic springs significantly slow down the computationtime and therefore linear elastic springs were used to simplify connectionsthat do not play a large role in the lateral behaviour of the building. More-over, the elastic springs were designed to remain elastic and therefore do notrequire the nonlinear design. Furthermore, the spring properties changedbased on the global coordinates so two defined springs are required for the3D analytical model. When defining the springs it is important to have allthe data on the connection such that all degrees of freedom of the connectionare defined correctly. Several link/support types are available in SAP2000;however, the multilinear plastic behaviour was used to allow for the plasticbehaviour in the connection. SAP2000 provides three different hysteresistypes for the multilinear links: kinematic, takeda and pivot. The hysteresisof t-stub connections from experimental tests showed to be most similar tothe pivot hysteresis type. The model was developed by Dowell, Seible, andWilson (1998) for reinforced concrete members. However, the hysteresis asshown in Figure 5.10 quite accurately predicted the response of the t-stubconnection. The pivot model allowed the researchers to modify the curveusing the variables in the model to tweak the curve to better capture the be-haviour of the t-stub connection. A detailed discussion on the pivot methodand variables can be found in Dowell et al. (1998).655.2. Modeling of lateral and gravity loading resisting systemsFigure 5.10: Pivot hysteresis model ( c© 2013, Computers and Structures, bypermission)The multilinear plastic link with pivot hysteresis was calibrated in SAP2000using a similar loading protocol of that by Piluso and Rizzano (2008) toreplicate their results. The resulting pivot model parameters are shown inTable 5.5. Where α1, α2, β1 and β2 represent various points on the hys-teresis curve originally defined using reinforced concrete columns (Dowell etal., 1998). For this thesis, the pivot model parameters were defined experi-mentally in SAP2000 to best represent the t-stub behaviour. Moreover, theexperimental results and resulting calibrated link model are shown in Figure5.11.Table 5.5: Pivot hysteresis model parametersPivot model parameters α1 50 α2 10 β1 0.1 β2 0.7 η 10 665.2. Modeling of lateral and gravity loading resisting systems-400-300-200-10001002003000 5 10 15 20 25Force (kN) Displacement (mm) (a) (b)Figure 5.11: T-stub connection: a) experimental result ([Piluso and Rizzano,2008] by permission from publisher); b) pivot model calibration in SAP2000SAP2000 does not fail a nonlinear spring during analysis when the finalconnection strength is reached. Once the springs reach the final nonlinearpoint in the defined curve they continue on with the same strength ratherthan failing. In this thesis, however, under dynamic analysis the t-stubconnections do not reach the failure point. However, if pushover analysiswas conducted manual post-processing would be required to determine thefailure point of the system.5.2.4 FramesThe steel plates, columns and beams were assigned frame propertiesin SAP2000. Nonlinearity in the frame member is possible through framehinges in SAP2000; however, the steel-timber hybrid being analyzed does notrequire hinges as the steel columns and beams were not designed to transfermoments and resist the lateral force due to an earthquake. The steel platesin the core system were designed as part of the lateral system but do notrequire plastic hinges as the links account for the nonlinear behaviour of thet-stub connection and plate.Frame releases were used to not allow the beam-column connections totransfer moments. All the columns ends were set to transfer zero momentand the frames were continuous with the CLT floor panels.67Chapter 6Seismic design and behaviourIn the review of the seismic behaviour of the steel-timber core wall build-ing both static and dynamic loads were used in the analysis. Static loadswere applied to the structure in SAP2000 according to the design loads rec-ommended by the NBCC 2010 and consistent with the benchmark buildingplans. The forces obtained following the ESFP method were applied on thestructure to validate the derivation of deformation equation for the systemusing first principles.The dynamic loads were applied to the building to replicate an earth-quake event (time-history analysis) and therefore were computationally timedemanding. To reduce the computational effort fast nonlinear analysis(FNA) was used for the time-history analysis. FNA relies on the modalanalysis using Ritz vectors to reduce the complications of the equilibriumrelationship for the structure using the elastic structural system (CSI, 2013).The method essentially separates the nonlinear components from the elasticones to reduce computational time. The method works best with systemsthat have a limited number of nonlinear members that are mostly links. Forthese reasons, FNA was used in the analysis in the following sections.6.1 Equivalent static force procedureThis section outlines the design of the building using an equivalent staticforce based design methodology as recommended by the NBCC. Details ofthe loading and climatic information for the site can be seen in Chapter 5.The ductility (Rd) and overstrength (Ro) factors were taken as 2.0 and 1.5,respectively, based on recommendations by the CLT Handbook (Gagnonand Pirvu, 2011).686.1. Equivalent static force procedureThe initial design base shear on the building was found using the ESFPbase shear equation (Equation 1.1). The period was determined using theempirical formula for shear walls and the seismic weight was determinedbased on the gravity loading system as defined in Chapter 5. Initially, thedesign base shear was much larger due to the fundamental period empiri-cal equation (0.55 s). However, the code allows up to twice the empiricalperiod for shear wall systems if the analytical model shows a larger period.Eigenvalue analysis in SAP2000 showed that the fundamental period of thestructure was 0.85s. For this reason the design base shear was reduced to1635 kN.Using the deformation equations formulated in Chapter 3 the lateralsystem was designed by limiting the drift in the structure under the ESFPloading to 1.5%. The limiting drift was chosen as the structural layout doesnot allow for excessive deformations. Drift was assessed using SAP2000based off the validation in Chapter 4. The T-stub connectors plasticallydeform to provide the necessary ductility in the system. Details of thethree connections (A, B, and C) used in the design are shown in Figure 6.1where connection A represents the four bolt connection to the foundationand connections B and C represent the two bolt connections between floors.The high strength four bolt t-stub connection was necessary transfer thelarge axial loads to the foundation.696.1. Equivalent static force procedure01002003004005006007000 10 20 30 40 50Force (kN) Displacement (mm) ABCFigure 6.1: T-stub connections for 3D ESFPPlate size varied based on the t-stub connector used as the plate mustremain elastic during an earthquake event. Therefore, the plate size is alwayslarger than the t-stub flange thickness. Details of the timber-steel core wallsystem are shown in Table 6.1.706.1. Equivalent static force procedureTable 6.1: ESFP design detailsCLT core walls Floors Thickness (ply) 3 - 8 5 1 - 2 7 T-stub connections Floors Type 2 - 7 C 1 B 0 A Plate size Floors Thickness (mm) 6 - 8 12 4 - 5 16 2 - 3 20 1 32 The design was evaluated by applying bi-directional ground motions tothe building in SAP2000. Bi-directional ground motions were used to betterunderstand the building behaviour in a realistic 3D earthquake event.6.1.1 Ground motion selectionGround motions were selected using a model that is based on a multiple-conditional-mean-spectra (CMS) method (Atkinson and Goda, 2011; Baker,2011). The model was updated from the Geological Survey of Canada (GSC)model from the current NBCC to include recent earthquakes, magnitude-occurrence relations, new ground motion prediction equations, extendedsource model and adopts mean estimates rather than median estimates(Atkinson and Goda, 2011).The improved model was used to select the ground motions to apply onthe proposed building. As the building is designed to be in Vancouver thesite class was set to C. The first three fundamental periods are used by the716.1. Equivalent static force proceduremodel in addition to the site class to select and scale the ground motions.The first fundamental period represents the anchor for the ground motionselection process and the second and the first three periods are used to selectthe range of vibration period.Ten ground motions were selected with the CMS model (Atkinson andGoda, 2011) to design and validate the proposed timber-steel hybrid struc-ture. Eigenvalue analysis with SAP2000 showed that the first fundamentalperiod of the building was 0.85 seconds; therefore, the anchor period wasset as 0.75 seconds with a range from 0.25 to 1.0 second. The model definesthe earthquakes by magnitude, distance and earthquake type. This allowsfor seismic de-aggregation based on the earthquake type such as crustal,interface and inslab. Figure 6.2 displays the 2% in 50 years uniform hazardspectrum for the ten selected ground motions.00.20.40.60.811.21.40 1 2 3Acceleration (g) Period (s) Figure 6.2: 2% in 50 years uniform hazard spectrumThe selected ground motions show a good distribution of magnitude,distance and earthquake type that will be able to capture any problems withthe structure. The pseudo acceleration for the ten sets of earthquake groundmotions can be seen in Figure 6.3. One ground motion set represents two726.1. Equivalent static force procedureground motions for the two directions of loading necessary for bi-directionalNLTHA.00.511.522.50 1 2 3Pseudo acceleration (g) Period (s) Figure 6.3: Pseudo acceleration response spectra for the ten sets of groundmotions6.1.2 NLTHA on the 3D timber-steel structureFigure 6.4 outlines the NLTHA on the 3D structure. The ground motionsets (Set 1 (Figure 6.4a)) are applied on the analytical building (Figure 6.4b)and the results are then computed in post processing. Figure 6.4c shows thedrift of the building in the X and Y directions and Figure 6.4d shows theresponse of the top of the building in both the X and Y directions.736.1. Equivalent static force procedure-0.2-0.15-0.1-0.0500.050.10.150.20.250 50 100 150 200Acceleration (g) Time (s) X-directionY-direction0123456780 50 100 150 200 250 300 350Storey Drift (mm) X-directionY-direction-300-200-1000100200300-300 -200 -100 0 100 200 300Y-displacement (mm) X-displacement (mm) (a) (b)(d)(c)Figure 6.4: Bi-directional NLTHA on timber-steel hybrid structureThe results of the NLTHA are shown in Figures 6.5 to 6.7. The driftresults for the 3D building remain below the 1.5% drift limit for both X andY directions as shown in Figures 6.5 and 6.6, respectively. The interstoreydrift results also show good performance for the timber-steel core wall system(Figure 6.7 and 6.8).746.1. Equivalent static force procedure0123456780 100 200 300 400Storey Displacement (mm) Figure 6.5: Building drift results of NLTHA for 2% in 50 years in the X-direction0123456780 100 200 300 400Storey Displacement (mm) Figure 6.6: Building drift results of NLTHA for 2% in 50 years in the Y-direction756.1. Equivalent static force procedure0123456780 0.5 1 1.5 2 2.5Storey Interstorey drift (%) Figure 6.7: Interstorey drift results of NLTHA for 2% in 50 years in theX-direction0123456780 0.5 1 1.5 2 2.5Storey Interstorey drift (%) Figure 6.8: Interstorey drift results of NLTHA for 2% in 50 years in theY-direction766.2. Energy based designAn average of the interstorey drift results would not provide a realistichazard level for the earthquakes as two of the ten ground motions weresubstantially more damaging than the average. The result is two groundmotions that have interstorey drift values close to 2%. The NBCC allows upto 2.5% interstorey drift (National Research Council (NRC), 2010); however,due to the structural layout of the building a 2% limit was suggested. Eachearthquake shows a trend of changing interstorey drift from floors 7 to 8.This is due to the high roof not being a full floor but rather just a corner ofthe structure.6.2 Energy based designThe proposed methodology was modified from the work of Choi et al.(2006). The steel-timber core wall EBD methodology follows the flow chartas shown in Figure 6.9.776.2. Energy based design1. Select earthquake records 2. Determine target displacement (um) and target ductility ratio (μ) 3. Convert MDOF structure to SDOF structure 4. Run NLTHA on the SDOF structure 5. Determine fundamental period (T) 6. Compute the input energy (Ei) 7. Compute the yield base shear (Vy) and elastic energy (Ee) 8. Compute the plastic energy demand (Ep) 9. Distribute plastic energy based on shear distribution and design the lateral system 10. Tn = Tn1 yes no 11. Validate design with NLTHA on the designed MDOF structure Figure 6.9: EBD flow chart786.2. Energy based designStep 1: Select earthquake recordsEarthquake records are selected based on the location the building isto be built. These ground motions govern the design of the structure asthe response spectras are developed from this set of ground motions. Theground motions used in the ESFP study were used for this EBD study asthe target period and range of period is expected to be similar to the EBDperiod. The pseudo acceleration (Figure 6.3) and velocity spectra (Figure6.10) were developed using the program Bispec (Hachem, 2004). Bispec isa nonlinear spectral analysis software that uses earthquake ground motionrecords to perform uni-directional and bi-directional dynamic time historyanalysis on SDOF systems (Hachem, 2004).0501001502000 1 2 3Pseudo velocity (cm/s) Period (s) gm1 gm2 gm3 gm4 gm5gm6 gm7 gm8 gm9 gm10Figure 6.10: Pseudo velocity response spectraStep 2: Determine target displacement and ductility ratioThis EBD methodology relies on a target displacement (uT ) to computethe ductility ratio (µT ) based off the known yield displacement (uy) of thesystem. Target drift is based on the structure type and desired performancelevel (Ghobarah, 2001). The yield displacement is found for the steel-timber796.2. Energy based designhybrid system using the equations as described in Chapter 4 with no inelasticbehaviour from the connections. The target ductility ratio is then definedas the target to yield drift (Equation 6.1).µT =uTuy(6.1)For the timber-steel hybrid system the interstorey drift was limited to2% as this would result in an overall drift of 1.5% to meet the life safeperformance level (Ghobarah, 2001). Therefore, the yield drift of the systemwas found to be 143mm and the target drift was set as 369mm. These driftsresulted in a target ductility of 2.5.Step 3: Convert the MDOF structure to an equivalent SDOFstructureThe yield and target displacement are then used to derive the equivalentSDOF yield (uy,eq) and target displacement (uT,eq) (ATC, 1996):uT,eq =uTΓ1φt1(6.2)uy,eq =uyΓ1φt1(6.3)where Γ1 is the modal participation factor and φt1 is the fundamentalmode shape vectors roof storey component. The MDOF to equivalent SDOFtransformation for the timber-steel core wall 2D structure is shown in Figure6.11.806.2. Energy based designMuTSDOFMDOFkuT,eqFigure 6.11: MDOF to SDOF transformationStep 4: NLTHA on SDOF structureBased on constant ductility, NLTHA is carried out on a SDOF structurewith a bi-linear force displacement relationship. Damping is assumed tobe 5% of the critical damping. A period range of 0.01 to 3.0 s is used tocompute the acceleration, velocity and energy spectra. Again, this studyused the program Bispec (Hachem, 2004) to construct the spectra. Theinput energy for various target ductility is shown below in Figure . 6.12816.2. Energy based design020004000600080000 1 2 3Input energy / mass (cm2/s2) Period (s) 1246Figure 6.12: Input energy for various target ductilityStep 5: Determine the periodFor the first iteration of the methodology the empirical formula from theNBCC for shear walls is used (Equation 6.4). After the first iteration thefundamental period will be re-iterated and therefore this empirical formulais only a starting point and does not dictate the final design.T = 0.05(hn)3/4 (6.4)where hn is the height of the structure.Step 6: Compute the input energyThe input energy (Ei) is estimated using the energy-balance concept asdescribed in Chapter 1:826.2. Energy based designEi =12M1S2v =12M1(T1Sa2pi)2(6.5)where M1 is the first modal mass, Sv is the pseudo velocity and Sa is thepseudo acceleration. This study uses the pseudo acceleration to estimatethe input energy (Figure 10). The pseudo acceleration in determining theinput energy is found by taking the average pseudo acceleration of the 10ground motions. The average acceleration of the ten ground motion recordsis shown below in Figure 6.13 along with the design iterations.00.10.20.30.40.50.60.70.80 1 2 3Acceleration (g) Period (s) Iteration #1 (T = 0.55 s, 0.631g)  Iteration #2 (T = 0.967 s, 0.352g) Final iteration (T = 1.139 s, 0.277g)  Figure 6.13: Average acceleration for each trial periodHowever, this estimate of input energy has shown to underestimate theearthquake energy input to the structure (Choi et al., 2006). Therefore,Choi et al. (2006) recommend a modification factor (α) to estimate thecorrect input energy (E∗i ):836.2. Energy based designα =(Ei,inelasticEi,elastic)·(VeqSv)2(6.6)where (Ei,inelastic/Ei,elastic) is the ratio of inelastic to elastic input energyfor the target ductility and Veq is the equivalent velocity:Veq =√2Eim(6.7)The equivalent velocity can be plotted using the above equation and thepseudo velocity as shown in Figure 6.14. Furthermore, the inelastic to elasticinput energy ratio is plotted in Figure 6.15.00.511.522.533.50 1 2 3Veq / Sv Period (s) Iteration #1 (T = 0.55 s, 2.04)  Iteration #2 (T = 0.967 s, 1.99)  Final iteration (T = 1.139 s, 1.87)  Figure 6.14: Ratio of the equivalent velocity to pseudo velocity846.2. Energy based design00.511.522.50 1 2 3E i,inelastic / Ei,elastic Period (s) Iteration #1 (T = 0.55 s, 1.02)  Iteration #2 (T = 0.967 s, 0.869)  Final iteration (T = 1.139 s, 0.765)  Figure 6.15: Ratio of inelastic to elastic input energyThe modification fator (α) is then multiplied by the input energy (Ei)to obtain the correct input energy (E∗i ):E∗i = αEi (6.8)Step 7: Compute the yield base shear and elastic energyThe yield base shear and elastic energy (Ee) are then determined:Vy =E∗iuTeq(1− 12µT) (6.9)Ee =12uy,eqVy(6.10)856.2. Energy based designStep 8: Compute the plastic energy demandFinally, the plastic energy demand for the inelastic system (E∗p) is esti-mated:E∗p = β(E∗i − Ee)(6.11)where β is the correction factor to account for the overestimation ofhysteretic energy ratio (Choi et al., 2006):β =Eh/EiEp/Ei(6.12)where Eh is the hysteretic energy and found through the NLTHA in step5. The ratio of hysteretic to input energy is plotted below in Figure 6.16.00.10.20.30.40.50.60 1 2 3E h / Ei Period (s) Iteration #1 (T = 0.55 s, 0.539)  Iteration #2 (T = 0.967 s, 0.503)  Final iteration (T = 1.139 s, 0.498)  Figure 6.16: Hysteretic to input energy ratioThe plastic energy demand must now be modified for the MDOF struc-866.2. Energy based designture:E∗pM = γE∗p (6.13)where γ is the ratio of plastic energy for a MDOF to an equivalent SDOF:γ =Ep,MDOFEp,SDOF(6.14)Step 9: Distribute energy based on shear distributionThe energy distribution was determined for the proposed system accord-ing to the t-stub connections. Therefore, the axial distribution was the bestrepresentative for the dissipation of energy. The ESFP designed 2D struc-ture is analyzed using NLTHA to determine the energy distribution. Theresults for each earthquake are averaged to determine the final distribution(Figure 6.17).876.2. Energy based design123456780 0.2 0.4 0.6 0.8 1Storey Normalized axial connection demand Figure 6.17: Normalized axial distribution ratioThe plastic energy demand from step 8 is then applied to the structureaccording to the distribution and the connections are designed to dissipatethis energy. Limiting factors on the connection design are the resultinginterstorey drift and maximum displacement. The t-stub connections usedin the 2D design are shown in Figure 6.18.886.2. Energy based design01002003004005006000 10 20 30 40 50Force (kN) Displacement (mm) DEFGHFigure 6.18: T-stub connections for 2D EBDStep 10: Fundamental period checkUsing eigenvalue analysis in SAP2000 modal analysis is carried out todetermine the fundamental period of the designed structure. If the newperiod from SAP2000 is the same as the assumed period from step 5 thenthe EBD methodology is complete. If the periods are not the same, thenew period is used as the input for step 5 and steps 6-10 are performedagain. These iterations continue until the fundamental periods converge.The design iterations for the 2D structure are shown in Table 6.2 below.896.2. Energy based designTable 6.2: EBD process for steel-timber hybrid structure with 1.5% targetdriftTrial 1 2 ∙∙∙ Final Period (s) 0.550 0.967 ∙∙∙ 1.139 Acceleration (g) 0.631 0.352 ∙∙∙ 0.277 Ei (kN∙mm) 902.8 874.9 ∙∙∙ 740.1 Veq / Sv 2.040 1.990 ∙∙∙ 1.870 Ei,inelastic / Ei,elastic 1.020 0.869 ∙∙∙ 0.765 Ei* (kN∙mm) 3832.0 3010.7 ∙∙∙ 1979.8 Vy (kN) 20.715 16.400 ∙∙∙ 10.516 Eh / Ei 0.539 0.503 ∙∙∙ 0.498 (Eh / Ei) / (Ep / Ei) 0.710 0.663 ∙∙∙ 0.656 Ep,MDOF / Ep,SDOF 0.759 0.759 ∙∙∙ 0.759 EpM*(kN∙mm) 1563.8 1146.6 ∙∙∙ 746.5 T-stub7 G G ∙∙∙ H T-stub6 G G ∙∙∙ H T-stub5 G G ∙∙∙ H T-stub4 G G ∙∙∙ H T-stub3 G G ∙∙∙ H T-stub2 F G ∙∙∙ H T-stub1 D E ∙∙∙ E The final EBD design for the 2D structure is shown in Table 6.3.906.2. Energy based designTable 6.3: Final EBD design detailsCLT core walls Floors Thickness (ply) 1 - 8 5 T-stub connections Floors Type 1 - 7 H 0 E Plate size Floors Thickness (mm) 3 - 8 12 2 16 1 20 Step 11: Validate the designThe final step in the methodology is to validate the EBD by applying theselected ground motions from step 1 on the MDOF structure. The resultsshow how the building behaves under the earthquake load and whether thetarget displacement is met for the system. Figure 6.19 shows the interstoreydrift results (solid) with the ESFP results (dotted). Moreover, Figure 6.20shows the maximum displacement results for the EBD methodology on thetimber-steel hybrid structure.916.2. Energy based design0123456780 0.5 1 1.5 2 2.5Storey Interstorey drift (%) Figure 6.19: Interstorey drift results for EBD0123456780 100 200 300 400Storey Displacement (mm) Figure 6.20: Maximum displacement for EBDResults show that the EBD performed as expected with an interstoreydrift value less than 2%. Moreover, the target drift was not exceeded for926.3. Summarythe earthquakes. Therefore, the performance level desired was achieved.However, under the proposed EBD methodology some of the earthquakesshould have exceeded the target drift as the average acceleration values areless than the maximum ground motions. There are a variety of ways tocalculate the input energy modification factor for the actual system. Thisresearch used the same method as Choi et al. (2006) suggested for BRBframed structures. Modifying this equation for a much lower ductility systemwill give more suitable results. When comparing to the work completed byChoi et al. (2006) for BRB structures the design method should be alteredas the BRB members are responsible in dissipating all the lateral forces inthe system. The proposed system has many components contributing to thelateral resistance and therefore the t-stub connections do not have the samelarge effect on the system as the BRB members.6.3 SummaryBoth the ESFP and EBD results prove that the proposed timber-steelcore wall system is feasible for high seismic regions. The interstorey driftof the ESFP designed building under bi-directional loading remained under2% and showed a good drift response in both X and Y directions.An EBD methodology for the proposed structure was developed basedon research by Choi et al. (2006). The methodology allows for the user todefine a target performance level and reach that performance level throughseveral iterations based on the fundamental period. The resulting EBDdesign for the 2D structure shows good behaviour under the earthquakeload. Target drift was not exceeded and the interstorey drift remained below2%. Furthermore, the design method allowed for appropriate selection ofconnections based on energy demand.The comparison between design methodologies (Figure 6.19) displayedthat the EBD structure had slightly larger interstorey drift values thanthe ESFP structure. Large differences were not expected between designmethodologies as the force modification factors used in the ESFP design andthe target ductility used in the EBD methodology were similar. The similar936.3. Summaryoverall stiffness of the two gravity designed systems further contributed tothe similar results.94Chapter 7Conclusions and futurerecommendations7.1 Summary of findingsThis thesis examines a novel timber-steel core wall system for high seis-mic regions. The timber-steel core wall system utilizes CLT panels contin-uously connected to steel plates secured together with t-stub connections.The proposed system couples the light and stiff CLT panels with the ductilebehaviour of the steel t-stub connections. The t-stub connections test be-haviour was replicated analytically using a pivot hysteretic model (Piluso etal., 2001; Piluso amd Rizzano, 2008). The bracket connections were respon-sible to transfer shear and axial loads from the diaphragm to the panels.The brackets were elastically modelled in the analytical models accordingto a study by Gavric et al. (Gavric et al., 2014). The proposed hybridtimber-steel core walls were derived using first principles and validated withSAP2000. Results of the validation showed the timber-steel core walls werecapable of resisting large overturning moments from seismic loads. Thecapacity of the system is adjusted by modifying the plate and t-stub con-nections and thickness of CLT panel. Confirming that the analytical modelwas behaving as designed; the model was used to analyze both the 2D and3D performance of the structure under different design methodologies.First, the system was designed following the NBCC’s ESFP accordingto the concrete benchmark building plan, climatic data and occupancy. Thegravity frame system was designed first according to the largest load com-bination following the NBCC (National Research Council (NRC), 2010).957.1. Summary of findingsOnce the columns, CLT floor and beams had been sized the building wasdesigned following the ESFP. The design results are discussed and show thatthe timber-steel core walls are capable of resisting the high seismic loads inthe analysis. The ESFP design was then analyzed using 10 bi-directionalground motions based on a conditional mean spectrum developed by Atkin-son and Goda (Atkinson amd Goda, 2011). Results of the study show thatthe proposed timber-steel hybrid structure stays under the 2% interstoreydrift limit. The interstorey drift values were as expected except for thehigh roof that saw varying increases and decreases in interstorey drift. Thisinconsistent result was observed to be a result of the smaller floorplan ofthe high roof floor (Figure 1.1). The maximum drift of the designed build-ing remained under 1.5% for all 10 bi-directional ground motions furthervalidating the proposed timber-steel core walls.Using the same gravity system the lateral system was designed using theproposed PBSD EBD methodology. PBSD was used on this system to definethe target drift before design and achieve an efficient design based on the sitespecific ground motions and structural system rather than code developedempirical formulas and force modification factors. EBD was used to utilizethe energy balance relationship to design the structure based on iterationsfor the fundamental period. The hybrid building was simplified to a 2Dframe for EBD with one of the six bays infilled with the timber-steel corewall system (Figure 6.11). The 2D structure was designed for a target inter-storey drift of 2% therefore allowing a building drift of 1.5% as observed bythe ESFP 3D building response. The design iterations, iterations and finaldesign results are presented and discussed. The design was then validatedby applying the 10 ground motions on the 2D frame. The results showedthat the building remained under the 1.5% drift limit and the imposed 2%interstorey drift limit. When compared to the ESFP designed 2D behaviourthe results show the benefits of designing following the EBD approach. Bydesigning the t-stub connections to dissipate the plastic energy based on theaxial distribution the interstorey drift results become much more desirableas the connections are appropriately assigned.967.2. Conclusions7.2 ConclusionsThe proposed timber-steel core wall system performed well in resistingthe seismic forces from the earthquake ground motions on the benchmarkbuilding studied. The benchmark building plan irregularity had a largeimpact on the performance as the cores were located in opposite corners ofthe building. This irregularity resulted in the second two mode shapes beingcontrolled by torsion. This structural layout restricted the performance ofthe building when compared to a building with core walls located on thesame center line or with core walls in the center of the building. With lessirregularity in the building layout the developed core wall system could bedesigned to deform more resulting in a smaller earthquake load due to thelarger displacement ductility. The core wall system performed well underthe benchmark building plan but has the ability to perform better with aless irregular layout as the t-stub connections can provide much more lateraldisplacement through plastic dissipation.In the force based design a ductility of 2.0 was used as recommendedby the CLT handbook. This was a conservative ductility as the handbookductility was derived using a different structural system with less plasticdeformation capacity. The value was chosen as no other study had beendone on the timber-steel core wall system. The results show that a ductil-ity of 2.0 was indeed conservative. Realistically, a larger ductility is moresuitable for the proposed system. The ductility would be even larger if thebuilding layout was less irregular as previously stated. This is one of themain reasons PBSD was studied as the method does not rely on generalseismic modification factors. There is just one ductility and overstrengthvalue for a structural system in the NBCC. These modification factors donot change based on the structural layout or height. With PBSD the de-signer defines the target displacement and the ductility is determined basedon the structural system, height and layout.The t-stub connections were designed between each floor even thoughthe upper floors had no plastic demand. As Figure 4.8 showed, the axial de-mand is very small in the upper floors when subjected to extreme earthquake977.3. Future recommendationsevents. Other connection details such as conventional CLT connections dis-cussed in Chapter 3 could be used in the upper floors to save money andproduce a more efficient structure. The effect damping has on the structurewas not explicitly studied and several assumptions were made regarding theeffect is has.7.3 Future recommendationsThe following list outlines some aspects of the timber-steel core wallsystem and the proposed EBD methodology that need further study in orderto ensure design application and feasibility.− Validate the equations derived in Chapter 4 through experimental testson the timber-steel core wall system.− Use finite element software that better captures the nonlinear be-haviour of the t-stub connection. SAP2000 has limited hysteresistypes; other software programs allow the user to define the hysteresisthat would better dictate the nonlinear cyclic behaviour of the t-stubconnection.− Consider various CLT panel configurations within the timber core wallsystem. Additional ductility could be introduced to the core systemthrough friction or connections within the timber core wall system.− Investigate the fire performance of the core system and modify thesystem to protect the steel elements in the CLT system.− Refine the EBD methodology. By extending the proposed EBD method-ology to 3D it would better capture the behaviour of the timber-steelcore system and allow for a more robust design.− Investigate alternative options for estimating the correct input energy(E∗i ). Other modification factors may better estimate the input energyto the proposed system.987.3. Future recommendations− Test various layouts of the proposed core system. The building studiedis irregular in shape, by considering a more high-rise style layout withone core wall area and a smaller floor plan the strength and ductilityof the system could be tested to its full extent.99BibliographyActon Ostry Architects. (2016). Brock Commons Phase 1. 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